FOUNDATION DESIGN
Content
University of Nairobi
Department of Civil and Construction Engineering
Geotechnical Engineering (FCE 511)
Teaching notes
By Sixtus Kinyua Mwea
2015
University of Nairobi –FCE 511 Geotechnical Engineering IV
Syllabus
FCE 511 - Geotechnical Engineering III
Foundations:
Shallow Foundations
Introduction. Foundation loading intensities. Bearing capacity, (ultimate, safe, gross and
allowable). Influence of ground water table, sloping ground, inclined and eccentric loads on
allowable bearing capacity. Design of shallow foundations for shear strength and settlements.
Examples of foundation design (e.g. strips, pad), combined footings, raft footings.
Piled Foundation
Types of piles driven and bored pile, friction and end bearing pile. Design of piles by soil
mechanics methods, end bearing, skin friction and ultimate bearing resistance. Piles in sands.
Piles in cohesive soils - total and effective stress analysis. Design from pile tests data.
End bearing piles on rock. Settlement of piles. Dynamic formula. Negative skin friction. Pile
groups - bearing capacity in cohesive and cohesionless soils.
Introduction to Earth Dams
Design of earth embankment - homogenous and zoned dams. Definitions e.g. fetch, water
spread, shell free board etc. Factors influencing site selection. Spillways. Settlements of
embankments. Protection of upstream and downstream slopes.
Site Investigations
Introduction, purpose of Site Investigation, organization of Site investigation for different types
of structures e.g. buildings, irrigation or water supply projects, highways and airport pavements,
etc. Methods of Investigation. Sampling. Borehole logs. Geophysical methods. Geotechnical
reports.
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Table of contents
Syllabus ................................................................................................................................... i
Chapter one ................................................................................................................................ 1
Shallow foundations ................................................................................................................... 1
1.1
Types of foundations .................................................................................................. 1
1.2
Introduction to shallow foundations ........................................................................... 1
1.2
Bearing capacity of soils ............................................................................................ 2
1.2.1
Bearing capacity terms ......................................................................................... 2
1.2.2
Ultimate bearing capacity..................................................................................... 3
1.2.3
The net foundation pressure ............................................................................... 12
1.2.4
Allowable bearing pressure ................................................................................ 13
1.2.5
Field methods for the determination of bearing capacity of soils ...................... 14
1.2.6
Presumed bearing capacity of soils and rocks .................................................... 23
1.3
Proportioning of shallow foundations ...................................................................... 24
1.3.1
Contact pressure distribution .............................................................................. 24
1.3.1
Proportioning the foundations ............................................................................ 25
1.3.2
General consideration in the selection of the foundation depth ......................... 34
1.3.3
Foundations for common buildings.................................................................... 35
1.4
Foundations for difficult soils .................................................................................. 36
1.4.1
Foundations on expansive clays ......................................................................... 36
1.4.2
Foundations on loose sands ................................................................................ 41
1.5
Tutorial examples on chapter one ............................................................................ 43
Chapter two .............................................................................................................................. 45
Deep Foundations ..................................................................................................................... 45
2.1
Pile foundations ........................................................................................................ 45
2.1.1
Introduction ........................................................................................................ 45
2.1.2
Classification of Piles by materials and construction......................................... 46
2.1.3
Driven piles ........................................................................................................ 48
2.1.4
Bored piles.......................................................................................................... 51
2.1.5
Determination of pile load carrying capacity ..................................................... 53
2.1.6
Determination of load carrying capacity dynamic methods............................... 59
2.1.6
Determination of load carrying capacity pile testing ......................................... 61
2.1.7
Negative skin friction ......................................................................................... 62
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2.1.8
2.2
Pile groups .......................................................................................................... 64
Drilled piers and Caisson Foundations..................................................................... 66
2.2.1
Drilled piers ........................................................................................................ 66
2.2.2
Caisson Foundations .......................................................................................... 66
2.4
Examples of Piling Schemes .................................................................................... 71
2.5
Tutorial examples on chapter two ............................................................................ 71
Chapter Three ........................................................................................................................... 73
Introduction to Earth Dams ...................................................................................................... 73
3.1
Introduction .............................................................................................................. 73
3.2
Selection of type of earth dam .................................................................................. 74
3.2.1
Diaphragm types ................................................................................................ 74
3.2.2
Homogenous types ............................................................................................. 75
Zoned types ..................................................................................................................... 75
3.2
Design Principles ..................................................................................................... 76
3.3.1
Foundation design .............................................................................................. 76
3.3.2
Embankment Design .......................................................................................... 79
3.3
Inspection of existing dams ...................................................................................... 81
3.4
Examples of earth dams in Kenya ............................................................................ 82
Chapter Four ............................................................................................................................. 88
Site Investigation ...................................................................................................................... 88
4.1
Introduction .............................................................................................................. 88
4.1.2
4.2
Planning a site investigation ............................................................................... 89
Preliminary and detailed stage site investigations .................................................... 91
4.2.1
Preliminary stage site investigations .................................................................. 91
4.2.2
Detailed stage site investigations ....................................................................... 92
4.2.3
Sampling............................................................................................................. 97
4.2.4
Scope of Site Investigation ............................................................................... 101
4.2.5
Site Investigation Reports ................................................................................ 102
References: ............................................................................................................................. 103
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Geotechnical Engineering IV
Week
Introduction
1 2
3
4
5
6
7
8
9
10 11 12 13 14 15
Shallow foundations
Foundation intensities
Bearing Capacity
Factors that influece bearing capacity
Design of shallow foundations
Piled foundations
Types of piles
Types of piles
Driven piles
Bored piles
Pile load capacity
Settlement of piles
Negative skin friction
Pile groups
Introduction to earth dams
Definitions (fetch water
freeboard)
Design of earth embankment
Site selection
Spillways
Settlement of embankments
Protection of slopes
spread,
Continuous Assessment Test
Site investigation
Introduction
Purpose of site investigation
Organization of site investigation
SI for different schemes
Methods for site investigation
Geotechnical reports
Revision and tutorials
Main examinations
Target dates
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Chapter one: - Shallow foundations
1.1
Types of foundations
Foundations that are encountered in practice may be classified into two broad categories
namely shallow and deep foundations. Under shallow foundations the following categories
are usually encountered:a)
b)
Strip foundations for wall and closely spaced columns
Spread or isolated footings for individual columns. In this category it is usual
to consider combined foundations for two or three closely spaced columns as
spread or isolated footings
c)
Raft foundations covering large sections of the foundation area
The design and construction of shallow foundations is dealt with in this chapter.
Under deep foundations the following two types of foundations are encountered :a)
b)
Piles
Caissons
The design and construction of deep foundations is dealt with in the next chapter.
In the selection of the foundations to adopt for a structure it is usually necessary to
consider the function of the structure, its loads, the subsurface conditions and the cost of the
foundation being adopted in comparison to other possible types of foundations.
1.2
Introduction to shallow foundations
The foundation is the part of the structure that transmits the loads directly to the underlying
soil. If the soil is sufficiently strong it is possible to use shallow foundation. On the other
hand if the soil is not strong enough the foundation is taken deeper into the ground and is
referred to as a deep foundation. A definition which sometimes conflicts with the definition
of the shallow foundation defines a shallow foundation as one whose depth is less or equal to
its least width. The foundation must satisfy two fundamental requirements:-
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Shallow foundations
1. The factor of safety against shear failure must be adequate. A value of 3 to 5 is
usually specified.
2. The settlement of the foundation should be tolerable and in particular differential
settlement should not cause any unacceptable damage o interfere with the function of
the structure.
3. The allowable bearing capacity is defined as the pressure which may be applied to the
soil to enable the two fundamental conditions to be satisfied
The damage being mitigated in the design of the structures can be classified as architectural,
functional or structural. In the case of framed structures settlement damage is usually
confined to the cladding and finishes (architectural damage). It is usual to expect a certain
amount of damage. What is critical is to ensure that the damage to the services is limited.
Angular distortion limits were proposed by Craig (1987) and are shown on Table 1.1. In
general the limiting angular distortion to prevent damage is 1/300. For individual footings
this translates to a maximum settlement of about 50mm in sand and 75mm in clay. An
accurate damage criterion is to limit the tensile strain at which the cracking occurs. The
concept of tensile strain should be used in analysis using an idealization of the structure and
the foundation in elastic strain analysis when the fundamental properties of the foundations
are known.
Table 1.1 Angular distortion limits
1/150
1/250
1/300
1/500
1/600
1/750
Structural damage of general buildings may be expected
Tilting of high rigid buildings may be visible
Cracks in panel walls expected
Difficulties with overhead cranes
Limit for buildings in which cracking is not permissible
Overstressing of structural frames without diagonals
Difficulties with machinery sensitive to settlement
The design of the foundations is usually a two process exercise. The first is to determine the
allowable bearing of the soil while the second is to size the foundation on the design strata
based on the allowable bearing capacity. These two parts are now discussed.
1.2
Bearing capacity of soils
1.2.1 Bearing capacity terms
The following terms are used in bearing capacity problems
Ultimate bearing capacity is the value of the average contact pressure between the foundation
and the soil which will produce shear failure in the soil.
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Shallow foundations
The net foundation pressure is the increase in the pressure at the foundation level due to the
structure loads
The safe net foundation pressure is the net foundation pressure divided by a suitable factor of
safety
Allowable bearing pressure is the maximum allowable net loading intensity on the soil
allowing for both shear and settlement effects.
1.2.2 Ultimate bearing capacity
If a load is increased at the foundation level, shear failure would take place in the foundation
at a load which can be referred to as failure load. The resulting pressure at the base of the
foundation is known as the ultimate bearing capacity of the soil
Three distinct modes of failure have been identified and these are illustrated in
Figure 1.1 in the case of strip footing. As the pressure increases on the foundation layer the
state of plastic equilibrium is reached initially in the soil around the edges of the footing and
then spreads downwards and outwards. Ultimately the state of plastic equilibrium is reached
throughout above the failure surfaces. The soil around the footing heaves on both sides. At
the moment of failure one side continues to settle at a higher rate and the strip footing tilts.
This behavior is exhibited by soils of low compressibility (Figure 1.1a). . Local shear failure
is characterized by local development of plastic conditions usually below the foundation. The
plastic conditions do not reach the surface and only slight heaving is expected. This kind of
failure is expected with soils of high compressibility and is associated with large settlements
(Figure 1.1b). These soils include dense and stiff soils. Punching shear occurs when shearing
takes place directly below the footing under compression from load. No heaving is of the
ground is expected by the side of the footing. Large settlements are characteristics of this
mode of failure and are typical of soils of high compressibility and foundations at
considerable depth (Figure 1.1c). In general the mode of failure will depend of the
compressibility of the soil and the depth of the foundation.
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a)
Shallow foundations
General shear failure
Pressure
a
Settlement
b)
local shear failure
c
b
c) Punching shear failure
Figure 1.1 Modes of failure of foundations
Bearing capacity by use of earth pressure analogy
The earth pressure analogy can be explained by consideration of a strip footing on a
cohesionless soil as shown on Figure 1.3
q
p
γD
Figure 1. 2 Pressure below a strip footing
The vertical pressure is q which is a result of the structure loads. By use of Rankine active
pressure theory, a lateral pressure p holds the soil in equilibrium below the foundation. For
particles just beyond the edge of the foundation the lateral pressure is more than the vertical
pressure γD resulting from the overburden. The vertical pressure γD is the minor principle
stress and p is the principal stress. By use of the Rankine earth pressure theory Equations 1.1
through 1.3 can be deduced.
p q(1sin)/(1sin) (inside the foundation)
1. 1
p D(1sin) /(1sin) (outside the foundation)
q D((1sin)/(1sin))2 (ultimate bearing capacity) 1. 3
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Shallow foundations
For a cohesionless soil the bearing capacity is dependent on the overburden and equals to
zero for a foundation on the ground surface. Bells development for a c-υ is given in Equation
1.4
q D((1si)/(1sin))2 2c ((1sin)/(1sin)3 2c (1sin)/(1sin)
.1. 4
For a purely υ =0 soil the ultimate bearing capacity is given by Equation 1.5
q D 4c
1. 5
Bearing capacity by use slip circle analogy
The slip circle analogy can be explained by consideration of a strip footing on a cohesive soil
as shown on Figure 1.3
B
q
D
O
B
πB
Figure 1. 3 A slip circle analogy of a strip footing
The foundation is assumed to fail by rotation about a slip surface of radius equal to the width
of the base B and at the edge of the foundation O. At ultimate conditions the disturbing
moment (Md) is given by Equation 1.6
Md q*L*B* B
2
1. 6
The resisting moment (Mr) about O is a summation of the resistance due to the cohesion on
the cylindrical surface, on the vertical surface and the weight of the overburden as given in
Equation 1.7
Mr cLB2 CDLB DLB
2
2
1. 7
At ultimate conditions the disturbing moment is equal to the resisting moment and the
ultimate bearing Equation for a υ = 0 soil is given by Equation 1.8
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Shallow foundations
q 6.28c(1 0.32D 0.16D)
B
c
1. 8
Plastic theory failure
A suitable failure under a strip footing is shown on Figure 1.2. The footing of width b and
infinite length carries a uniform pressure of magnitude qf. The shear strength parameters for
the soil are c and υ. The unit weight of the soil is assumed to be zero. At ultimate bearing
capacity the soil is pushed downwards into the soil mass producing a state of plastic
equilibrium in the form of an active Rankine zone below the footing where the angles ABC
and BAC are each 45+υ/2. The zone ABC resists movement and is intact with the base. It
suffers no much deformation. The downward movement of the wedge ABC forces the
adjoining soil to move sideways. Passive Rankine zones ADE and GBF are developed and
angles AEF and BFG are 45-υ/2. these zones confine the movement of the wedge EDA and
BGF. The transition between the downward movement of the wedge ABC and the lateral
movement of the wedge EDA and BGF takes place through zones of radial shear ACD and
BCG. The surfaces DC and CG are logarithmic spirals. The soil above EDCGF is in a state
of plastic equilibrium while the rest of the soil is in state of elastic equilibrium.
qf
qo
A
B
45+φ/2
F
45-φ/2
E
D
C
G
Figure 1. 4 Failure under a strip footing
Using plastic theory the ultimate bearing capacity below a strip footing on a surface of a
weightless soil is given by Equation 1.9. This is for undrained condition where υu = 0
qf (2)cu 5.14cu
1.9
In general the foundation is located at a depth and imposes a surcharge qo = γD. The weight
of the surcharge and the pressure of the foundation produce stresses on the moving masses of
soil at plastic conditions.
The ultimate bearing capacity of the soil under shallow strip footing can be
expressed by the following general equation suggested by Terzaghi.
qf 0.5BN CNc DNq
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Shallow foundations
Nγ, Nc and Nq are bearing capacity factors which depend on the values of υ. Nγ
represents the contribution to the bearing capacity resulting from the self weight of the soil.
Nc is the contribution due to the constant component of the shear strength and Nq is the
contribution of the surcharge pressure. Values of Nγ, Nc and Nq can be obtained from
Equations 1.11 through 1.13 the values for Nc and Nq were suggested by Meyerhof (1955)
while the values of Nγ, were suggested by Hansen (1970) These values are plotted in terms of
υ in Figure 1. 5.
Nc (Nq 1)cot
Nq tan2 (45/ 2)e tan
N 1.5(Nq 1)tan
1. 11
1. 12
1. 13
Values of Nc, Nq, Nγ
Nq
100
Nγ
Nc
10
1
0
10
20
30
40
φ - Degrees
Figure 1. 5 Bearing capacity factors for shallow foundations
Bearing capacity for square, round and rectangular foundations
The problem involves extending what is basically a two dimension problem in a strip footing
to a three dimension problem in other foundation shapes. The bearing capacity factors for
square and round foundations are shown on Equations 1.14 and 1.15 respectively.
q 0.4BN 1.3cNc DNq
q 0.3BN 1.3cNc DNq
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1. 14
1. 15
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Shallow foundations
The factors for rectangular footing are an interpolation of the square and the strip footing and
are shown on Equation 1.16
q 0.5BN (10.2B/ L) cNc (10.3B/ L) DNq
1. 16
It showed be noted that the values of the bearing capacity factors are very sensitive to the
values of shear strength parameters c and υ. Due consideration should therefore be given to
the degree of accuracy of these values. In general the following observations have been made
a)
b)
c)
d)
e)
In cohesive soils the contribution of cohesion c to the bearing capacity dominates
The depth factor dominates for cohesionless soils
The base factor is usually neglected for values of B less than 4 meters
A footing at the surface has no bearing capacity if Nγ is neglected
The equations are applicable to uniform soils and in the case of stratified soils an
engineering judgment is always required.
Skempton’s values of Nc
Skempton (1951) showed that for a cohesive soil (υ =0) the value of Nc increases with the
value of foundation depth D. He suggested that the values of Nc applicable to circular, square
and strip foundations are given in Figure 1.6. The value of the rectangular footings of
dimensions BxL (where B<L) is the value of a square footing multiplied by (0.84+0.16B/L).
10
9
Nc
8
7
Nc (Strip)
6
Nc (Circular or
Square)
5
4
0
1
2
3
D/B
Figure 1.6 Skempton’s values of Nc for a φ =0 soil
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Shallow foundations
Eccentric and Inclined loading
Eccentric and inclined loadings have an effect of reducing the foundation bearing capacity.
In the case of a foundation with a vertical load such that the eccentricities are e b and el
(Figure 1.7 ) the effective foundation dimensions are shown as B’ and L’ The resulting load
is distributed over the effective foundation dimensions. The values of B’ and L’ are given in
Equations 1.17 and 1.18
B' B2eB
L' L2eL
1. 17
1. 18
In the case of inclined load (Figure 1.8) on a width B and inclination the effective
foundation width is B-2e. In addition the bearing capacity factors are multiplied by the
inclination factors shown on Equations 1.19 and 1.20
ic iq (1/90o )2
i (1/)2
1)
1. 19
1. 20
The base of a long retaining wall is 3m wide and is 1m below the ground in front
of the retaining wall. The water table is well below the base level. The vertical
and horizontal components of the base are 282kN/m and 102kN/m respectively.
The eccentricity of the base reaction is 0.36m. The appropriate shear strength
parameters are c’= 0 and υ’ = 35o. The unit weight of soil is 18kN/m3.
Determine the factor of safety against shear failure
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Shallow foundations
282kN/m
1m
282kN/
1.14m
102kN/m
.36m m1.5m
Solution
For υ’ = 35o, Nγ = 41 and Nq = 33
The angle of the inclination to the vertical α = tan -1 (102/282) = 20o hence the inclination
factors according to Meyerhof are
The ultimate bearing capacity is given by
The factor of safety
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Shallow foundations
An alternative approach in the case of inclined loads is to use the empirical formula shown on
Equation 1.21
L’
Y
B’
B
eB
X
eL
L
Figure 1. 7 Effective dimensions for pads subjected to eccentric loads
PV / Pav PH / Pah 1
1. 21
Where Pv is the vertical component of the inclined load and PH is the horizontal
component of the inclined load. Pva is the allowable vertical load and PHa is the horizontal
load (a fraction of the available passive resistance).
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Shallow foundations
α
Pv
P
PH
e
B
Figure 1.8. Foundation with inclined load
1.2.3 The net foundation pressure
The actual pressure on the soil due to the weight of the structure is called the total foundation
pressure q. The net foundation pressure qnet is the increase in the pressure at the foundation
level. This is the total foundation pressure less the effective weight of the soil permanently
removed during excavation and is given in Equation 1.22
qnett q D
1. 22
For a strip footing the net foundation pressure is shown on equation 1.23
qnett 0.5BN cNc D(Nq1)
1. 23
The safe net bearing pressure (qsafe) is the net bearing pressure factored by an appropriate
factor of safety as shown on Equation 1.24
qsafeqnett/ FOS
1. 24
It is usual to use conservative factors of F usually between 3 and 5. Due to uncertainties in
the determination of the strength parameters
and determination of the of the service load,
for comparison the following factors of safety (Table 1.2) are used in other geotechnical
works
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Shallow foundations
Table 1. 2 Typical factor of safety values for geotechnical works
Failure mode
Shear
Shear
Shear
Seepage
Shear
Type of works
Earthworks
Retaining walls
Sheet piles
Uplift
Bearing Capacity
FOS
1.2-1.6
1.5-2.0
1.2-1.5
1.5-2.5
3-5
Effect of ground water
Water table below the foundation level
If the water table is at a depth not less than B below the foundation level the expression for
the net ultimate bearing capacity is given in Equation 1.23 above. However the when the
water table rises to a depth less than B below the foundation level Equation 1.25 is
applicable.
qnet CNc D(Nq1) subBN
1. 25
For cohesive soils the value of υ is small and the term ɣsubBNɣ is of little account.
Consequently the bearing capacity is not affected by the ground water variation below the
foundation level. For sandy soils the term CNc is zero and the term 0.5γsubBNγ is about half
0.5γBNγ. The effect of the groundwater is significant.
Water table above the foundation level
For this case the net ultimate bearing capacity is given by Equation 1.26. It is seen both
cohesive and cohesionless soils are affected by the water table rising above the foundation
levels
qnet CNc p'o (Nq1) subBN
1. 26
Where p’o is the effective overburden above the foundation level.
1.2.4 Allowable bearing pressure
In design, the settlement due to the safe net bearing pressure is computed. If the resulting
settlement is not acceptable then the pressure used in the determination of the settlement is
reduced. At the point when the settlement is acceptable then the pressure obtained is the
allowable bearing capacity of the soil.
In design the ultimate loads are obtained from structural analysis. The ultimate load
is converted into the service load. The gross load is the structural load above the ground floor
plus the overburden. The net load at the foundation level is the load at the ground floor in
addition to the weight of the foundation less any soil which has been replaced. For practical
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Shallow foundations
considerations it is therefore not necessary to consider the weight of the foundation below the
ground level (Figure 1.9)
P
Gross load =P + overburden
Pnet = P + foundation load – replaced
soil
=P
Figure 1. 9 Net load applied at the foundation level
1.2.5 Field methods for the determination of bearing capacity of soils
Plate bearing test
The test is particularly suited for the design of foundations or footings where it is considered
that the mass characteristics would differ from the laboratory tests and where the precise
values of settlement are required. The plate load test covers the determination of vertical
deformation and strength characteristics of soil insitu. From the data recorded the allowable
bearing capacity of the soil is estimated.
In the test an excavation is made to the expected foundation level. The plate usually
300 to 750 mm square should be rigid and flat. It is loaded by means of kentledge. The
kentledge can be any form of dead load including water, concrete blocks etc or tension piles.
The loading procedure can be either constant rate of loading or incremental loading
procedure as described below:Constant rate of penetration
This test is best suited to undrained conditions. In the test the load is applied in a controlled
manner to enable a continuous and uniform rate of penetration. The load is continued until a
penetration of 15% of the plate width is achieved. The ultimate load is considered to be the
load corresponding to the 15% of the plate width penetration.
Incremental load test
This test is best suited to drained conditions. In the test the load an estimate of the maximum
load is calculated. Five equally spaced increments are then selected. The load at each
increment is recorded together with the corresponding settlement. A load is maintained until
the penetration has ceased or when the primary consolidation is complete. The ultimate load
is considered to be the load corresponding to the 15% of the plate width penetration as in the
case of the constant rate of penetration test.
Plate bearing capacity test results
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Shallow foundations
The plate bearing test results are best reported in graphical way as shown on Figure 1.10.
The weak soft clays and loose sands will reach the ultimate bearing capacity in the region of
100-200 kN/m2 while the stiffer clays and dense sands and gravels will continue increasing
in the bearing pressure with increasing settlement.
Bearing Pressure (kN/m2)
0
200
400
600
800
1000
0
Settlement (mm)
-2.5
-5
-7.5
-10
Stiff clay, dense sand or gravel
Soft clay or loose sand
-12.5
-15
Figure 1.10 Typical plate loading test results
Estimation of allowable bearing pressure from plate bearing test results
The test is reliable only if the stratum being tested is reasonably uniform over the significant
depth of the full scale foundation. A weak stratum below the significant depth of the plate
but within significant depth of the foundation would have no influence over the plate test
results but will have a significant effect over the performance of the foundation (Figure 1.11).
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Shallow foundations
B
b
1.5b
1.5B
Weak stratum
Figure 1.11 Influence of weak stratum
Settlement of the stratum increases with increasing loaded area and the main
problem is in the extrapolation of the test results to full scale scenario. Ideally the plate test
should be carried out using plates of different sizes and at different depths. However, this is
usually not economical.
Notwithstanding this shortcomings the following procedure was been proposed by
Terzaghi and Peck (1948) and can be used as a guide to use of plate bearing test results. The
settlement of a square footing kept at a constant pressure increases as the footing size
increases. The relationship is shown on Equation 1.27 relates the settlement of the test plate
of 300 mm square and that of a square foundation of width B.
S S1 *(2B/(B0.3))2
1. 27
Where S1 = settlement of the loaded area under a 305mm plate for a given pressure
intensity p
S= the settlement of a square foundation of width B in metres under pressure p
In order to use the plate bearing results the maximum allowable settlement is determined. A
value of 25mm is generally accepted as an allowable settlement. S is then equated to 25 and
a numerical value of B is inserted in the formula to enable the determination of the S1. From
the relationship of p and s1 the value of p corresponding to the calculated value of S1 is the
allowable bearing pressure subject to any adjustments certain to the ground water conditions.
Standard penetration test
The test covers the determination of the resistance of soils particularly sand and loose to
medium loose gravels at the base of a borehole to the penetration of a split barrel sampler
when dynamically driven in a standard manner. In addition to the determination of resistance
the split sampler is used to obtain disturbed samples for determination of remolded properties
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Shallow foundations
namely particle size analysis and Atterberg limits when the sample has some degree of
plasticity. When used in gravels the sampler is replaced with a 60 o cone which does not
sample the soil.
Figure 1.12 shows the main features of standard penetration test equipment. The
drive shoe and the sampler consist of 51 mm external diameter and 35 mm internal diameter.
It is 450-600 mm long. This is connected to a drive assembly at the bottom of the boring
rods. A pick and release mechanism which ensures a free fall of a hammer weigh 65
kilograms through of 760 mm + or – 20 mm is used to drive the sampler or the cone in the
case of the gravelly strata
Figure 1.12 Standard penetration equipment.
The procedure requires that the borehole is cleaned carefully to ensure that disturbed soil at
the bottom of the borehole is removed. When boring below the ground water table it is
prudent to maintain the water in the borehole at the same level or higher than the general
ground water. A hydraulic balance is needed to avoid the risk of boiling of the strata at
bottom occasioned by a high hydraulic gradient.
The sampler and the hammer are lowered to the bottom of the borehole. If after
touching the bottom the sampler penetration exceeds 450 mm on its own weight and that of
the hammer, the SPT value also known as N value is recorded as zero. Otherwise after the
initial penetration on own weight the test is driven in two stages known as seating drive and
test drive
The seating drive is the initial 150mm penetration or 25 blows whichever is reached
first. The test drive is the further penetration of 300mm or 50 blows which ever is reached
first. The number of blows for the 300 mm penetration is recorded as the SPT value ‘N’. If
300 mm penetration can not be reached in 50 blows the test drive is terminated. In this case a
hard stratum has been encountered and further driving results in damage of the split sampler.
It is usual to record the blows for every 75 mm penetration. If the test drive is terminated the
penetration corresponding to 25 and 50 blows is recorded.
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Shallow foundations
Interpretation of Standard Penetration Test Results
The pore water pressure generated by the hammer during testing affects the value of N.
When the test is carried out below the water table in fine sand or fine silt the resistance
increases as a result increased pore water pressure which does not dissipate immediately. If
the measured N if greater than 15 a correction as shown on Equation 1.28 is performed.
TrueN151/ 2(N 15)
1. 28
The relative density of a soil affects the N values. Terzaghi and Peck (1948) evolved a
qualitative relationship between the relative density and the standard penetration N values.
Gibbs and Hortz put values of relative density. Table 1.3 shows the two relationships
Table 1. 3 Relationship of N values and the relative density of sands
N value
0-4
4-10
10-30
30-50
50+
Terzaghi and Peck
Very loose
Loose
Medium
Dense
Very Dense
Gibbs and Hortz
0-15%
15-35%
35-65%
65-85%
85-100%
The effective stress at the level of the test also affects the penetration of the SPT split barrel
sampler. This effect can be related to the effective overburden at the level of the testing.
Craig (1986) has summarized the correction of the overburden into Equation 1.29.
N' CN N
1. 29
Where N’=the corrected value of SPT
N=the measured value of the SPT or the true N in the case of the saturated loose
sands and silts
CN=Overburden factor
The relationship of CN and effective overburden is shown on Figure 1.12
University of Nairobi –FCE 511 Geotechnical Engineering IV
Effective Overburden (kN/m2)
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Shallow foundations
500
400
300
200
100
0
0
0.5
1
1.5
2
Correction factor CN
Figure 1.12 Estimation of N’ from measured values of N (Craig 1986)
Standard penetration resistance increases with increasing particle size, increasing overconsolidation ratio and increasing angle of internal friction of the soil. A correlation between
shear strength parameter and N, and effective overburden is shown on Figure 1. It provides
rough estimate of value of and should not be used for very shallow foundations.
SPT - N
50
40
φ=25
30
φ=30
φ=35
20
φ=40
φ=45
10
φ=50
0
0
50
100
150
200
Effective overburden (kN/m2)
250
Figure 1.13 Correlation between shear strength parameter φ and N and effective
overburden
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Shallow foundations
Estimation of allowable bearing pressure from standard penetration tests
In 1948 Terzaghi and Peck presented a chart as shown on Figure 1.14 for the estimation of
allowable bearing capacity while limiting the settlement to 25mm and differential settlement
to 75% of the maximum settlement. The procedure involves determination of the average
value of N’ from all the boreholes at the foundation level. The allowable bearing capacity for
the widest foundation is determined and then applied to all the foundations. Terzaghi and
Peck based his chart on foundations on unsaturated soils when the water table is at lower than
1.0B below the foundation. Thus when the water table is at 1.0B the reduction of the
allowable bearing capacity is zero. The reduction increases linearly as the water rises. When
the water table is at the ground level the reduction is 50%. Thus the provisional value of
allowable bearing capacity obtained from Figure 1.14 should be reduced by the factor Cw
shown on Equation 1.30
Cw 0.50.5Dw /(D B)
1. 30
Where Dw= depth of the water table below the ground level and D
D =the depth of the foundation
B = the width of the foundation
Figure 1. 14 Relationship between N and allowable bearing pressure
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Shallow foundations
Static cone penetration test
The test apparatus consists of a 60o cone as shown on Figure 1. The cone is subjected to
continuous penetration in the soil the rate of 15-20 mm per second. The tip has electrical
sensors for continuous recording of the resistance and penetration as shown on Figure 1. On
the more sophisticated cones the friction along the cone can be measured. In addition the
water pore pressure can also be measured. At every penetration the resistance is measured as
load/cone area and is plotted against penetration
Figure 1.15 Static cone penetration apparatus
Resistance = load/end area = qc (kN/m2)
0
200
400
600
0
Penetration (mm)
-2.5
-5
-7.5
-10
-12.5
-15
Figure 1. 16 Static cone penetration test results
University of Nairobi –FCE 511 Geotechnical Engineering IV
800
1000
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Shallow foundations
From the data the value of Nγ used in Terzaghi Equation for the Ultimate Bearing Capacity
can be estimated from Equation 1.31 the value of internal angle of friction can be obtained
from Figure 1.5 which then enables the determination Nq and the ultimate bearing capacity.
Other empirical values of qa can be obtained from equations 1.32 through 1.33
N qc /80
qa qc / 30 for B< 1.2m
qa qc /50*((B0.3)/ B)2 for B> 1.2m
1. 31
1. 32
1. 33
Allowable bearing capacity on rock stratum
The bearing capacity of rock is the highest that an engineer can expect to get. In some cases
the intact rock has unconfined compressive strength larger than the strength of the concrete
which goes to the making of the foundation. In this case it is the structural design of the
materials rather than the strength of the rock control the foundation design.
For ordinary structures when site investigation is performed by boring, bedrock need
be proved to a depth of three meters to discount the possibility of isolated boulders (Craig,
1987). When un-weathered rock has been reached in foundation construction, the allowable
bearing pressure is based on the inherent strength or the parent rock. The influence of joints,
discontinuities and shear zones is to reduce the allowable bearing capacity. The rock quality
designation (RQD) defined as the ratio of the total length of core of full diameter and length
greater than 100mm or greater to the length of the core run measures the extent of defects and
has been used in the determination of the allowable bearing pressure as shown on Table 1.
Table 1. Allowable bearing capacity RQD
RQD
100
90
75
50
25
0
Allowable bearing capacity (kN/m2)
29,300
19,500
11,700
6,800
2,900
1,000
Source Peck et al, 1973
Bowles (1982) stated that the settlement is more often the concern than the bearing capacity.
Consequently most effort should be taken in the determination of modulus E and Poisson’s
ratio η so that an estimate of the settlement can be made. Alternatively he suggested that one
should use a large factor of safety on the unconfined compression strength of the intact
fragments obtained from the borings. The factor of safety should depend on the RQD and
typically range between 6and 10.
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Shallow foundations
Tomlinson and Boorman (1986) reported the presumed bearing capacity must not
exceed half of the unconfined compression strength of the intact rock fragments. Ibi (1986)
reported presumed allowable bearing capacity values of various rocks varying from 12,500
kN/m2 for igneous and limestone rocks to as low as 150 kN/m2 for weak un-cemented
mudstones.
Rock strength designations based on the unconfined compressive strengths have
been suggested by BS 5930 (Ibi (1986) and the Canadian Geotechnical Society (Franklin and
Dussealt, 1989) are shown on Tables 1.4 and 1.5 respectively.
Table 1. 4 Rock strength designation by BS 5930
Classification
UCS (kN/m2
x103)
Very
Weak
Under 2
Weak
1.25 to
6
Mod
Weak
5 to
20
Mod
Strong
12.5 to
60
Strong
50 to
200
Very
Extremely
strong
strong
100 to Over 200
200
Source – Tomlinson and Boorman (1986)
Table 1. 5 Rock strength designation by Canadian Geotechnical Association
Classification
UCS (kN/m2
x103)
Extremely
Weak
Under
2
Very
Weak
2 to
6
Weak
6 to
20
Medium
Strong
20 to
60
Very
strong
100 to
200
Extremely
strong
Over 200
Source: Franklin and Dussealt (1989)
1.2.6 Presumed bearing capacity of soils and rocks
It is common to use presumed bearing capacity of soils and rocks. The values used have been
derived after many years of testing and performance monitoring of existing structures. These
values are usually conservative do not consider the overburden above the foundation level.
They can be used as preliminary values for the very large structures where an accurate
bearing capacity at the foundation level is needed. In the case of smaller structures these
valued can be considered as final. Table 1.6 shows the presumed bearing capacity of soils as
suggested by BS8004 (1986), while Table 1.7 shows the presumed bearing capacity values
used in Kenya. It is to be noted that difficult soils such as expansive soils, loose sands and
silts and made up ground should be investigated all the time.
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Shallow foundations
Table 1. 6 Presumed allowable bearing vales (BS 8004: 1986)
Category Types of soils and rocks
Rocks
Non
cohesive
soils
Cohesive
soils
Strong igneous and gneissic rocks in sound
Strong limestone and strong sandstones
Schists and slates
Strong shales, mudstones and siltstones
Dense gravel, or dense sand and gravel
Medium dense gravel or medium dense sand
and gravel
Loose gravel or loose sand and gravel
Compact sand
Medium dense sand
Loose sand
Very stiff boulder clay and hard clays
Stiff clays
Firm clays
Soft clays and silts
Very soft clays and expansive clays and silts
Peat, organic soils, made up ground and fill areas
Value
( kN/m2)
10000
4000
3000
>600
<200-600
<200
>300
100-300
<100
300-600
150-300
75-150
<75
Remarks
The foundations
should be taken to
un-weathered rock
The
foundation
width to be not
less than 1m and
water level not
less than below
the
foundation
level
Soils susceptible
to
long-term
consolidation and
settlement
Not applicable
Not applicable
Table 1. 7 Presumed allowable bearing values in Kenya
Soil and rock
Red coffee soil (Red silty clay)
Medium dense sand
Loose gravel (Murram)
Dense gravel
Compact gravel and weathered rock
Un-weathered rock
Expansive soils, loose sands and silts
1.3
Value (kN/m2)
80-120
100-300
100-150
200-400
350-600
>600
Not Applicable
Proportioning of shallow foundations
1.3.1 Contact pressure distribution
This is the distribution of the pressure below the base of the foundation and the ground. The
pattern of the distribution varies according to the stiffness of the foundation. The stiffness
may be described as yielding (elastic), rigid or flexible
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Shallow foundations
Yielding foundation
The stiffness of such foundation is zero. Here the contact pressure distribution has the same
variation as that of the load. Because of its zero stiffness there will be no moments induced
in the footing. Such a condition exists in fresh concrete before it sets. It has no practical
significance.
Rigid foundations
Contrary to the yielding foundation the rigid foundation has infinity rigidity. They are so
rigid that they do not deflect. Most of the foundations considered in practice are rigid
foundations. The analysis is simple and leads to economical design of the footings.
Flexible foundations
The stiffness of such foundations lies between rigid and the yielding foundations. The
foundations in this category deflect to a certain degree depending on the magnitude of their
stiffness. The analysis of such foundations is complicated but leads to an economical design.
However this is not usually done in practice and is not considered in these notes.
1.3.1 Proportioning the foundations
The proportioning of the foundations is usually the final step in the design of a structure. The
type of foundation, sizes and the level of the foundation depend on the result of the site
investigation. Usually partial factors would have been used in the design of the columns.
However unfactored loads would be used in the proportioning of the foundations. The
factored loads are however required in the determination of the foundation depths and design
of the foundation in accordance with BS 8110 (1997). The general procedure for the design
of the foundations follows the following steps
a)
Evaluate the allowable bearing pressure in a site investigation exercise
b)
Examine the existing and future levels around the structure and take into account the
ground bearing strata and the ground water level to determine the final depth of the
foundation
c)
Calculate the loads and the moments if any on the individual footings with partial
safety factors on the structural loads.
d)
Recalculate the loads and the moments on the individual columns and the walls
without partial factors of safety. In many cases it is sufficiently accurate to divide the
factored loads and moments with 1.45.
e)
Calculate the plan area of the foundation using unfactored loads
The plan area of the foundations is determined assuming that all the forces are
transmitted to the soils without exceeding the allowable bearing pressure. The distribution of
the pressure is assumed to be planar. In no case should the extreme pressure be less than
zero. All parts of the foundation in contact with the soil should be included in the assessment
of the contact pressure. Subsequently the designer carries out the structural design of the
foundations. Typical foundations are now discussed
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Shallow foundations
Strip and rectangular footings
A strip footing is significantly greater in length than in width. This type of foundation is used
to support walls and closely spaced columns. When and individual column is supported by a
footing then this foundation is referred to as a pad footing. When two or more columns are
supported by one footing, this is referred to as a combined footing.
Axially loaded strip and rectangular foundations
The contact pressure of these foundations is considered as uniform when loaded axially. The
pressure under the foundations should not exceed the allowable bearing pressure of the
supporting soil. Figure 1.17 shows the pressure distribution of such foundations.
a) Pad foundation
b) Strip foundation
d) Pressure distribution
c) Combined foundation
Figure 1.17 Pressure distribution below individual and strip foundations under axial
load
Eccentrically loaded rectangular foundations
When foundations are subjected to axial and moments at their foundations the soil pressure
resultant does not coincide with the centroid of the footing. The resulting pressure is a
combination of the compression and the moment stresses. While the columns can in almost
all cases resist the moments it is doubtful that the spread footing can sustain an applied
column moment. The base usually will rotate and induce more moment at the far end of the
column.
In conventional analysis the contact pressure distribution under eccentrically loaded
rectangular foundations (Figure 1.) are derived from the common flexural formula. The
general formula for the estimation of the pressure when there is eccentricity in the y and x
axis is given in Equation 1.34.
(x,y) PAMy Iy *xMx Ix *y
1. 34
Where
σ(x,y) = contact pressure at any given point (x, y)
P = the vertical load
x,y = coordinate of the point at which the contact pressure is calculated
My and Mx = the moment about y and x axis respectfully
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Shallow foundations
Ix and Iy = moment of inertia of the footing area about the x and y axis respectively
=L*B3/12 and BL3/12 respectively.
y
L
Mx and My
P
B
y ex
Mx
ey
x
ex
ex
My
Figure 1. 18 Rectangular foundation eccentrically loaded in two axis
When Equation 1.34 results in negative values in some areas, this means that the foundation
soil is taking tension. It is then necessary to change the dimensions to have only compression
pressure at the base. This is difficult and requires trial and error approach for solution of
maximum and minimum pressures. It is prudent to place the foundation such that that there
is only eccentricity in one axis direction as explained below.
Eccentrically loaded rectangular foundations in one axis
In design it is common to determine the magnitude of the contact pressure at the edges.
Equation 1.34 reduces to equation 1.35 shown below and Figure 1.19 shows the pressure
distribution.
q PBL(16e L)
1. 35
P
P
P
M
L
M
M
e LL
L
L
b) e=l/6
c) e>l/6
e
a) e<l/6
Figure 1.19 Soil pressures below footing
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Shallow foundations
When the eccentricity inside mid-third of the base (Figure 1.19a,e<l/6) the computed
minimum pressure is positive soil pressure and the computed maximum pressure should not
exceed the allowable bearing pressure. At e=l/6 Figure 1.19b the minimum soil pressure q=0
and the footing is fully effective in bearing. This limit of eccentricity means that as long as
the eccentricity is less than l/6 also described as falling within the mid-third of the foundation
the entire footing is effective. When the eccentricity is large (Figure 1.19c) and e>l/6 the
computed minimum pressure is negative soil pressure. This is an indication of a tensile stress
between the soil and footing. This in not feasible and the soil pressure has to be evaluated
neglecting any soil tension. The eccentricity is said to be outside mid-third.
For eccentricity outside middle third with respect one axis the maximum soil pressure
redistributes itself since the base cannot take negative pressure. The distribution of pressure
is triangular and is shown on Figure 1.20. The equations applicable in this case can be
derived as follows:P
M
B
L
L’
L’/3
e=M/P
Figure 1. 20 Eccentrically loaded rectangular out of middle third
L' L e
3 2
and
P q (BL')
2
Solving the two equations to obtain the maximum soil pressure q, Equation 1. is obtained
q
2*P
3B(l / 2e)
1.36
Rectangular combined footings
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Shallow foundations
It may not be possible to place columns at the centre of spread footings if they are near the
property line, near mechanical equipment or irregularly spaced columns. Columns located
off center will result in a non uniform soil pressure. In order to avoid the non uniform soil
pressure, an alternative is to enlarge the footing and place one or more of the columns in the
same footing to enable the center of gravity of the columns loads to coincide with the center
of the footing (Figure 1. . The assumption here is that the footing is rigid. The column loads
are taken as point loads and distributed into the footing. The footings are statically
determinate for any number of columns. The column loads are known and the resulting
pressure is shown in equation 1.37
q P/ A
1. 37
P1
P2
variable
S
Figure 1. 21 Combined rectangular footing
Trapezoidal shaped footings
A trapezoidal shaped footing is required when a combined rectangular footing will not result
in uniform pressure. This is usually so when the space between the combined footings is
constricted as shown on Figure 1.22.
b
a
X’
L
Figure 1. 22 Trapezoidal footing
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Shallow foundations
From Figure 1.22 the position of the centre of area of the footing is x’. The centre of the area
is to coincide with the center of gravity of the loads from the two or more columns being
supported by the trapezoidal footing. The position of the base cannot be extended beyond the
length dimension L. L is therefore a known dimension. The value of the area of the
foundation is obtained from the allowable bearing pressure and the total column loads (
A P/ qa ). . The area of the base is shown in Equation 1.38 and the position of the centre
of the area is shown in Equation 1.39. The solution to the two equations leads to unique
values of a and b representing the dimensions of the trapezoidal footing.
A a b L
2
x1 L * 2a b
3 a b
1. 38
1. 39
From Equation 1.39 and Figure 1.22 it can be seen that the solution for a=0 is a triangular
footing and for a=b it is a rectangle. The solution for a trapezoid footing exists only for
Lx1 L
3 2
Strap or cantilever footings
A strap footing is designed to connect an eccentrically loaded column to an interior column
as shown on Figure 1.23. The strap is used to transmit the moment caused by eccentricity to
the interior column footing so that a uniform soil pressure applied to both footings. The strap
serves the same purpose as the interior portion of combined footing and is used in lieu of
combined rectangular or trapezoidal footing. Equations 1.40 through 1.43 are used to
proportion the footing dimensions. The value of eccentricity e is chosen arbitrary by the
designer. Unique solution of the strap footing is not always possible
R1 *S1 P1S
R1 P1 S1
S
R2 P1 P2 R1
L1 / 2 e x
R1 B1 *L1 *qa
1. 40
1. 41
1. 42
and
R2 B2 *L2 *qa
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1. 43
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S
P1
x
Shallow foundations
e
P2
L2
L1/2
S1
R1
R2
Figure 1. 23 Typical strap footing
Three basic considerations for strap footing design are:a)
b)
c)
The strap must be rigid (Istrap/Ifooting>2. This rigidity is necessary to avoid rotation of
the exterior footing.
The footing should be proportioned to approximately the same soil pressures and
avoidance of large differential settlements
The strap should be out of contact with the soil so that there are no soil reactions and
is weightless
A strap footing is to be considered only as a last option when other options would not work.
The extra labor involved in the forming of the deep beam and accompanying costs make it
only an attractive alternative when other options have been exhausted.
Raft foundations
A raft foundation is a large concrete slab used as a foundation of a several columns in several
lines. It may encompass the entire foundation area or only a portion. Raft foundations are
generally used to support storage tanks, several pieces of industrial equipment or high rise
buildings. Figure 1.24 shows some typical raft foundations
A raft foundation is used where the supporting soil has a low bearing capacity.
Traditionally the raft is adopted when pad and structural wall foundations cover over half the
area enclosed by the columns and the structural walls. However this should be evaluated on a
case by case basis since the raft foundations end up with negative moments and top and
bottom reinforcement. This arrangement could end up being more expensive than closely
spaced pads which require only bottom reinforcement.
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Shallow foundations
(a
(b
(c
)
)
)
(a) Flat slab; (b) Thickened under columns or beam slab (c) Basement walls as part of the raft or
cellular construction
Figure 1. 24 Common types of raft foundations
The advantages of the raft foundations over the other foundations include:a)
b)
c)
The effect of combining the column bases is increase in the bearing capacity of the
foundation. This is because the bearing capacity increases with the breadth of the
base.
The raft foundations bridge over the weak spots
They reduce settlement and are particularly suitable for structures sensitive to
settlement.
Raft foundations are usually designed as infinitely rigid in comparison to the supporting soil.
This assumption simplifies the pressure under the raft to a linearly distributed contact
pressure. The centroid of the contact pressure coincides with the line of action of the
resultant force of all the loads acting on the raft. Figure 1.25 shows the pressure distribution
and the resultant of the vertical loads.
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Shallow foundations
Resultant of column and wall loads
σmin
σmax
Resultant of soil pressure
Figure 1. 25 Linear pressure distribution below a rigid raft
A raft foundation is considered as rigid if the column spacing is less than 1.75/λ. λ is given by
Equation 1.44
Ks *b
4*Ec *I
1/ 4
1. 44
Where Ks = coefficient of sub-grade reaction
B = width of strip of the raft between centers of adjacent bays
Ec = modulus of elasticity of concrete
I = the moment of inertia of the strip of concrete
λ. = characteristic coefficient
Bowles (1982) suggests that the coefficient of subgrade reaction be estimated from Equation
1.45.
Ks 40*F*qa
1. 45
Where F = the factor of safety applied to the ultimate bearing capacity
qa = the allowable bearing capacity
Equation 1.44 is applicable when the column loads do not vary in magnitude by more than
20%. The column loads should also be uniformly spaced. The design of the raft follows the
following basic steps
a)
b)
c)
Compute the maximum column and wall loads
Determine the line of action of the resultant of all the loads
Determine the contact pressure distribution using Equation 1.46. Figure 1.26 shows
the arrangement of the columns and the eccentricities with respect to x and y axis.
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Shallow foundations
P*ey * y P*ex *x
A
Ix
Iy
(x,y) P
1. 46
Where ∑P=total loads on the raft
A = Total area of the raft
x, y =Coordinates of any point on the x and y axis passing through the centroid of
the raft
Ix and I y = moment of inertia of the area of the raft with respect to the x and y axis
respectively
ex and ey = the eccentricities of the resultant force in the x and y direction
It is conventional to obtain the pressures at the four corners and then interpolate in between to
enable the determination of moments and shears for the structural design of the raft
y
P1
My
PP2
2
ex
P3
ex
∑P
B
ey
P4
P5
Mx x
ey
P6
P9
P7
P8
L
Figure 1. 26 Raft foundation plan showing column loads
1.3.2 General consideration in the selection of the foundation depth
Once the geometry of the foundation of the foundation has been found, it is necessary to
determine an appropriate depth of the foundation. The following are general considerations
which the designers should take into consideration.
a)
Usually the foundation should be placed below the depth with minimum moisture
variation over the years. This eliminates the shrinkage and collapse effects of the
foundation soil. In this country a depth of between 1.0 and 1.5 metres is usually
sufficient.
University of Nairobi –FCE 511 Geotechnical Engineering IV
- 35b)
c)
d)
e)
f)
Shallow foundations
The foundation should be placed below top soil and below depths with roots of tress.
The roots are potential water paths which weaken the foundations.
The foundations should be sited with due consideration to existing nearby structures.
The exaction of the foundation in the vicinity of the existing structures could lead to
loss of lateral support of the neighboring structures.
Special attention should be taken to foundations supported on expansive soils and
those on loose sandy silts which are likely to be saturated during the lifetime of the
structure.
For water structures viz: - river bridges it is necessary to take extra care to ensure that
scouring of the foundation vicinity does not impair the safety of the foundation. It is
usual to use gabions in areas where scouring is likely to erode the foundations such as
downstream of box culverts and around abutments and pier foundations
It is preferable to place foundations at one level throughout. None the less if it is not
practical to have the foundations at one level, the change of level should be at one
plane. Sloping foundation levels should be completely avoided even if they are on
rock. There is a risk of the foundation sliding.
1.3.3 Foundations for common buildings
This section deals with foundations for ordinary common buildings. These are single and
double storied buildings with structural walls as the main form of support. The spans should
generally not be bigger than six metres. The buildings are generally on good bearing soils.
The bearing soils include red coffee soils, gravelly soils and firm sandy, gravelly clays. The
footing for these common buildings is shown on Figure 1.27. The 600 mm width is a
practical width which allows masons to maneuver into the trench.
200-150 mm thick
masonry wall
100mm slab with BRC no 65 at the top
face
200-150 DPC
Damp proof membrane
150 mm minimum drop
100-200 mm thick hardcore
dropasountonsd
A minimum of 1000 mm
depth of foundations
600mm wide x 200mm deep
mass concrete foundation
Figure 1. 27 Typical strip footing for an ordinary building
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Shallow foundations
The following are the general considerations in the usage of the standard footing.
a)
b)
c)
d)
1.4
No reinforcement is needed for strips where the load can be distributed through 45o.
The foundations should be excavated and the last 150mm excavation be finalized
when the concreting can be done without further delay. This minimizes the softening
of the foundation
The mass concrete is in mass concrete usually by volume batching to achieve grade
15 concrete. A ratio of 1:3:6 for cement sand and ballast respectively is generally
sufficient.
Reinforced concrete foundations are done for areas with concentrated loads. These are
usually column supports. Grade 25 concrete is the lowest class of concrete allowed in
the new BS 8110, but grade 20 of concrete can be considered.
Foundations on difficult soils
1.4.1 Foundations on expansive clays
Introduction
The problems associated with expansive soils arise as a result of alternate heaving and
shrinkage of the clays. These soils are typically black or grey and are referred to as black
cotton soils in this country. The cycle of expansion and shrinkage is a result of ability of the
clays to take in water and retain it in its clay structure. The water absorption leads to
expansion of the clay and causes strains in the foundation and the structures supported
thereupon. The strains eventually cause the cracks to appear on the walls. The result is
structural safety and aesthetics of the buildings are compromised
The clay minerals include montmorillonite, illite and kaolinite as discussed in FCE
311. The montmorillonite clay mineral is particularly prone to heaving and shrinkage. Soil
having more than 20% of montmorillonite are particularly prone to swelling problems
In addition to visual identification the expansive soils can be identified by assessing
the swell potential of the soils. This is done by conducting an odometer test which measures
the free swell and the swell pressure attained in an odometer when a sample held in an
odometer ring is kept at the same volume as swelling is induced by allowing the sample to
take in water. Some of the Nairobi black cotton soils have been found to have a swell
pressure of up to 350 kN/m2. Chen ( ) has related swell potential to plasticity index as shown
on Table 1.8. The following methods can be applied to mitigate damage control
a)
b)
c)
Moisture control
Soil stabilization
Structural measures
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Shallow foundations
Table 1.8 Relationship of swelling potential and plasticity
Swelling Potential
Low
Medium
High
Very High
Plasticity index (PI)
0-15
10-35
20-55
Over 55
Source (Chen, )
Moisture control
The main course of heave and shrinkage is the fluctuations of moisture under and around the
structures in question. Depending on the topographical, geological and weather conditions
the natural ground water fluctuates during the year. This seasonal fluctuation decreases with
depth. In some areas the depth to the fluation zone is as low as 1.5 meters. In other areas it
will be deeper going down to over three meters. In addition to the ground water fluctuation
the surface water from rains or bust pipes seeps into the foundations and course moisture
migration.
A satisfactory solution to the problem would to devise an economical way of
stabilizing the soil moisture under and around the structure. It does not matter whether the
moisture is maintained high or low in so far as it can be maintained throughout the year. An
effective procedure of achieving this is to provide a water tight apron of approximately one
metre round the building. A subsurface drain one metre round the building is provided with
augur holes provided at every 2 meters. The holes are filled with sand and interconnected at
the top. In effect the augur drain is and the impervious apron ensures that the moisture at the
foundation area remains the same. Figure 1. 28 shows such an arrangement of the drains for
ensures that the moisture content of the foundations remain the same
The subsurface drain is used to intercept the gravity flow, or; perched water of free
water to lower ground. It also arrests capillary moisture water movement. The subsurface
drain should be lend to a positive outlet. In general the ground surface around the building
should be graded so that surface water will flow away from the building foundations all h the
time.
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Shallow foundations
Positive drain to outfall
away from the building
Building
a)
Location of sand drain around a building
Ground floor with double
mesh A142
Masonry
Original ground
walling
level
2 meter wide water
Compacted granular material at
tight apron
high water content
Coarse sand drains
at 2 metre intervals
Expansive soil
b) Sand drain and apron detail
Figure 1. 28 Typical sand drain treatment of a building
Soil stabilization
Soil stabilization consists of one of the following operations
(a)
Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction.
(b)
Compaction control
(c)
Soil replacement
(d)
Chemical stabilization
Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction involves the
flooding of the site under consideration prior to construction. The soil would heave and the
potential danger of cracking is eliminated. Pre-wetting has been used with success when the
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Shallow foundations
active zones are not large. It is very difficult to saturate high plasticity clays. There is danger
that expansion of the clays could continue after the construction has taken place. This
procedure should be considered for stabilizing pavement or canal linings. In only rare cases
should the method be considered for use below ground floor slabs. Its application below
building foundations is risky and questionable.
Compaction control has been used in pavement construction. Expansive clays expand very
little when compacted at low densities and high moisture contents. But will expand
considerably when compacted to high densities at low moisture contents. The approach is to
compact swelling clays at moisture contents slightly above their natural moisture content for
good result. In this method it is not necessary to introduce large amounts of water into the
soil. Dry compaction of expansive soils was done along the Lodwar-Kakuma road.
Soil replacement is the simplest an easiest solution for slabs and footings founded on
expansive soils. The expansive foundation soils are replaced with non-heaving materials.
The method requires the selection of the replacement material and the depth to replacement.
In Nairobi the depth of the expansive black cotton soils is in the region of 1.0 to 1.5 metres.
In this case it has been found desirable to remove the entire expansive soil below buildings
and replace with suitable granular material. When the expansive soil is deeper building slabs
can be constructed above the compacted soil covering the expansive soil but the foundation
of main structure needs further consideration.
This method is particularly useful for the construction of highway pavement in a site
completely overlaid with expansive soils where the alternative to reroute the road is not
viable. In this case it the lower expansive soils are overlaid with the compacted replaced
material to a depth of 1.5 metres.
Chemical stabilization is the process of mixing additives like cement and lime to expansive
soil to alter its chemical structure and in the process retard its potential expansiveness. Lime
reduces the plasticity of the soil and hence its swelling potential. The amounts used range
from two to eight percent by weight. Cement on the other hand reduces the liquid limit,
plasticity and potential volume change. Stabilization has been used mainly in highway and
airport construction.
Structural measures include several methods have been reported in literature such methods
include
(a)
(b)
(c)
Floating foundation
Reinforcement of brick walls
Foundation on piles
Floating foundation concept is a providing a stiffened foundation. This is essentially a slab
on ground foundation with the main supporting beams resting on non-cohesive non heaving
material. The slabs are designed fixed on the beams that assuming a heave pressure of 20
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Shallow foundations
kN/m2. This magnitude is small considering that the swell pressure of the expansive soils
commonly found in Kenya has been estimated at between 300 and 500 kN/m2. Results of
such an approach have been mixed where they have been tried. This method needs further
research.
Reinforcement of brick walls have been tried in South Africa. In this method reinforcement is
placed in brick walls. The reinforcement is placed where cracking usually takes place. This
is typically above and below openings. The structure is made also semi flexible by providing
joints in the brickwork so that when heave takes place the building will conform to the new
ground shape and consequently reduce the bending moment induced in the walls. The joints
are typically 1.5cm.
Foundation on piles is a very successful procedure which ignores the heave by placing the
footing to a sufficient depth (Figure ). The depth of the pile should leave an expansion zone
between the ground and the building to allow the soil to swell without causing detrimental
effect to the building. One way of installing the piles is to provide a pile with bell at the
bottom. The bell or under reamed section should be well below the active zone. The bell is
installed with special equipment and anchors the pile into the ground. The pile can be
installed in an oversize shaft which is subsequently filled with straw saw dust as filler to
eliminate uplifting of the pile by heaving soil. Alternatively the pile could be a straight and
the effect of the uplift calculated using Equation 1.47 The friction below the active zone is
utilized in the calculation of the bearing capacity of the pile.
1. 47
Where
= the total uplift
D = the diameter of the pile
h = the depth of the pile in the active zone
u = the swelling pressure
f = the coefficient of friction between the pile and the soil
f may be taken as 0.15 while the swelling ;pressure varies between 250 and 500 kN/m2
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Shallow foundations
Straight pile
friction
Uplift
zone
zone
Skin
Active
Stable
friction
Uplift
Skin
h1
h
Beam
zone
Stable Active zone
Beam
Under ream
pile
Figure 1. 29 Pile systems for expansive soils
1.4.2 Foundations on loose sands
Foundations on loose sands are particularly difficult due to the likelihood of collapse in the
event of large storms. The storms result in the realignment of the sand particles and
consequent settlement due to repacking of the sand support. This has resulted in large cracks
in buildings which have been placed on this type of foundation soils. The foundation soils
subsequently loose there bearing capacity and the result is settlement of the foundations. The
superstructure has to absorb the settlement usually with resultant cracks of walls and
structural elements.
A real case story is one of the Garissa teachers college whose buildings were placed
on sand strata. The area is generally dry but when the rain comes, it usually very heavy and
comes in large storms. The performance of the three building types of structures adopted at
Garissa teachers college forms a case study whose findings are used to suggest a construction
procedure for foundations and masonry superstructures on loose sands.
The main teaching bungalow consisted of buildings constructed with a ground beam
which was framed with columns and a concrete roof slab. The masonry was thus reinforced
at the corners with columns and subsequently bound at he top by a ring beam and at the
bottom with a ground beam. These types of buildings were found to have performed well
several years after construction. This type of construction produced a satisfactory type of
constructed and when the buildings were inspected ten years after construction the structural
frames and the infill masonry walls were performing well.
The second type of buildings consisted of three and four and three storied flats. As in
the case of the previous buildings these types of buildings were found to have performed well
ten years after construction
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Shallow foundations
The third type of the buildings was the staff residential bungalows. These were
constructed with a ground beam and masonry walls. The roof of the buildings was a concrete
slab. However as the rains came and went in there stormy characteristics the residential
houses developed cracks in the walls. The cracks were particularly severe in the external
walls and after about 10 years of service and needed attention (Plate 1.1
Based on the satisfactory behavior of the framed structures it was found prudent to
introduce columns at the masonry wall corners in a repair scheme. Plate … It is therefore
recommended for foundations on loose sands the masonry should be reinforced with columns
at the corners. In addition the foundations should be kept as far as is possible free from
percolating water. In this way the in the event of settlement the frame will be able to absolve
the stressed attributable to additional settlement and reduce the severity of the cracks.
Plate 1.1 Cracks in the walls occasioned by settlement of the foundation
Plate 1.2 Introduction of columns to stiffen the walls
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Shallow foundations
Tutorial examples on chapter one
1) A footing 2.25 m square is located at a depth of 1.5m. The strength parameters are c’= 0
and υ’ = 38o. Determine the ultimate bearing capacity
a)
b)
If the water is is well below the foundation level.
If the water table is at the surface.
Given that the unit weight of sand above the water is 18 kN/m3. The saturated
unit weight of soil is 20kN/m3.
Ans – A 2,408kN/m3 B, 1,365 kN/m3
2)
A strip footing is to be designed to carry a load of 800kN/m rum at a depth of
0.7m in gravelly sand. The appropriate shear strength parameters are c’= 0 and υ’
= 40o. determine the width of a footing if the a factor of safety of 3 is specified
assuming that the water level may rise to the foundation level Above the water
table the weight of the gravelly sand is 17 kN/m 3. The saturated unit weight of
strata is 20kN/m3.
Ans – 1.55m
3)
A footing 2m square is located at a depth of 4 m in stiff clay of saturated unit
weight 21kN/m3. The undrained strength of the clay at a depth of 4m is cu=
120kN/m3 and υ’ = 0. For a factor of safety of 3 with respect to shear failure,
what load can be carried by the footing
Ans – 1680kN
4)
You are responsible for the design of a combined footing to support two columns
as shown in the figure below. The vertical dead loads on column A and B are 500
and 1400KN respectively. The design requires that the resultant of the column
loads acts through the centroid of the footing. In addition the dead loads, columns
A and B also can carry vertical live loads of up to 800 and 1200 KN respectively.
The live loads vary with time, and thus may be present some days and absent
other days. In addition the live load on each column is independent of that on the
other column. Check that the design meets all eccentricity requirements if the
worst possible combination of live loads is imposed
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Shallow foundations
5) A column is carrying a load of 1200kN. The column is located 300mm form the
boundary of wall. Calculate the pressure distribution if the column is founded on a
square base of 1500mm x1500mm. is the foundation safe if the allowable bearing
pressure is estimated at 300kN/m2
6) An internal column is carrying a load of 2400kN. It is located 3000mm from the
column described in Question 1 Design:a.
a suitable combined base for the two columns
b.
A suitable strap footing for the two columns
7) Your client acquires the next plot and you are not limited by the boundary wall.
Calculate the safe bearing pressure below the columns described in questions 1 and 2.
Assume a detailed site investigation has established the following strength parameters.
C’ = 10kN/m2, υ’ =20o, γsat = 18 kN/m2, γb= 16 kN/m2,
4
Four columns are carrying a tower. If the columns are on a square grid of
2.5mssquare, calculate the pressure at each of the four column positions if a raft
foundation of 3 mmx3m is designed to carry the foundation loads estimated at
4000kN, 5000kN, 6000kN and 7000kN
University of Nairobi –FCE 511 Geotechnical Engineering IV
Chapter two: Deep
Foundations
Deep foundation can be categorized into three major types. These include
i.
ii.
iii.
Pile foundations
Drilled piers
Caisson foundations.
The ground and structural conditions which require the use of the two types are discussed
under each of the sections dealing with the two types of the foundations.
2.1
Pile foundations
2.1.1 Introduction
Pile foundations are structural members used to transmit surface loads to lower levels in the
soil mass. They are used when soil beneath the level at an appropriate raft or conventional
footing is too weak or too compressible to provide adequate support to the structure load.
The piles have small cross-section area compared to their lengths. The pile materials
generally include timber, steel or concrete. The transfer is by vertical distribution of load
along the pile surface and at the pile end point.
Piles may be used in the following circumstances
a)
To transfer loads to a suitable bearing layer when weak strata is ignored and the load
is transferred to an overlying strong bedrock or compact layer.
b)
To transfer load through the shaft friction when compact layer is very deep and would
be impractical to reach it
c)
To support structures over water where conventional exaction and construction of the
foundation is not possible or very expensive to achieve.
d)
To reduce settlement and in particular differential settlement
e)
Based on cost. It might prove economical to drive piles down the strata and then
build on top of the piles instead of having to excavate deep layers and then construct
ordinary foundations
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- 46f)
g)
h)
Deep Foundations
In structures which have considerable uplift, horizontal and/or inclined forces. This is
especially true for marine and harbor works.
To increase the bearing capacity by vibration and compaction of granular layers of
soil.
In soils where deep excavations would result in damage of existing buildings.
Piles can be distinguished by the function they are intended to perform or by the material and
construction procedures used in their construction. The various types of piles by function are
shown on Figure 2.1. The main function of the piles is to take the loads by end bearing or by
friction or by combination of the two. Other functions exist and two which can be sited here
include tension piles and fender piles. The tension piles take lateral forces in place of
traditional retaining walls while fender piles also referred to as dolphin piles are marine
structures principally for taking horizontal loads from vessels in the docking areas. Section
2.2 is presentation of piles by their material and construction procedures.
Soft soil
Soft soil
Soft soil
Friction
Firm
resistance
strata
Hard
strata
End bearing pile
Friction pile
Combination
Impact from floating
object
Tension resistance
Tension pile
Dolphin or fender pile
Figure 2. 1 Types of piles by function
2.1.2 Classification of Piles by materials and construction
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Deep Foundations
Piles are constructed in a variety of properties of materials, construction methods and
functions. This makes as simple classification difficult. Notwithstanding theses difficulties
they are classified in accordance with the pile materials and method of construction (Figure
2.2). This classification also identifies the pile materials. The principal timber materials are
timber, concrete and steel.
Types of piles
Driven piles
Large displacement
Preformed. Solid
or hollow tubes
closed at the end
and left in position
Cast in place formed
by driving closed
tubular sections
and then filling the
void as the tube is
withdrawn
Solid
Pre-cast concrete or
Timber. Formed to
required lengths as
units with mechanical
a) H and
pipe piles
Bored piles
Small displacement
Replacement
Steel sections
H Piles
Open ended tubes
unless a plug forms
during driving
A void is formed
by excavation.
the void is filled
with concrete
sides may be
Supported or
unsupported
Hollow
Steel or concrete
tubes closed at the
bottom. Filled or
unfilled after driving
b) RC
Precast pile
The supporting may effected permanently
by casing or
Temporarily by casing or drilling mud
(Betonite) or
By soil on a continuous auger
c) Shell
Pile
Figure 2.2 Principal Types of piles
University of Nairobi –FCE 511 Geotechnical Engineering IV
d) Cast in-situ
tube withdrawn
e) Bored pile
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Deep Foundations
2.1.3 Driven piles
To install prefabricated and some form of cast in place piles it is necessary to displace soil by
driving the piles. The piling is commonly done by means of a hammer. The hammer
operates between guides or leads by use of lifting cranes. The leads are carried by the cranes
such that they can drive vertical or raking piles. The piling assembly may be mounted on
base suitable for operation on land or on a floating pontoon in the case of piling in the sea.
The hammers may be free falling operated by a clutch release mechanism.
Alternatively they are powered by diesel or steam. There are several forms of mechanical
devices and equipment in the market used by piling contractors. In order to reduce the
impact stresses on the hammer and the piles it is normal to strike the pile through a hammer
cushion. The elements of cushion vary but are mainly wood packing in a steel cap or dolly.
The various elements in the cushion not only protect the top of the pile but have a significant
influence on the stress waves developed in the pile during the driving. The rating of a
hammer is based on the gross energy per blow. For a drop hammer the rated energy is the
product of the hammer and the height of fall. The efficiency of the hammer is the defined as
the energy delivered at impact divided by the gross rated energy. Energy having been lost in
the dropping of the hammer to pile. For driving piles to great length the hammers have
energies of between of between 50kNm to over 180kNm.
Piles are installed by impact hammers and driven to a resistance measured by
number of blows required in the final stages of piling. For wood piles the energy would be
limited to about 3 to 4 blows per inch when energy of 15kNm is applied by the hammer. If
the pile is to be driven through heaving strata then, it might be necessary to predrill the
borehole where the pile is to be driven. This eliminates undesirable heaving. Additionally if
the pile is to be driven through dense layers of sand and gravel it is possible to loosen the
hard strata by sending a stream of water jet with specially adapted equipment. The various
types of driven piles are now described.
Timber Piles
Timber piles are made of trunks of timber. The timber should be preserved to prevent decay.
Untreated timber embedded below the ground water table has a long life. If the timber is
exposed to alternating wetting and drying it is subject to decay. These types of piles are not
very common.
Steel Piles
Steel piles (Figure 2.2a) are usually in form of H-Piles and pipe piles. H piles are preferred
where high depth is required while the pipe piles are usually filled with concrete after driving.
In the case of H-Piles the flanges and the web are equal thickness in order to
withstand large impact forces. Steel H piles penetrate the ground more readily than other pile
types because of the relatively small cross-section area. They are subsequently used to reach
stronger bearing stratum at great depth. Steel H piles have also relatively large bearing
capacity of between 500 and 2,000 kN per pile depending on the size of the H section. The
pile H sections are usually 250x250 to 350x350 with varying section thickness.
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Deep Foundations
Pipe piles are of the range of 250mm to 750 mm diameter. The wall thickness is
usually over 2.54mm. In the event that the wall thickness is less than 4.54mm the pile has to
driven with a mandrel. When the thickness of wall is over 2.5mm the pipe acts with any
concrete in carrying the load. Pipe piles are usually driven with the lower end closed with a
plate. In some instances conical driving shoes have been attached. The advantage is not
significant.
Steel piles are subjected to corrosion. The corrosion is minimal when the entire pile
is embedded in natural soil. However, the corrosion can significantly increase in the event of
entrapped oxygen. Zones of water table variation are particularly vulnerable. Severe attacks
are encountered on sea structural sections exposed to high and low water tides where the salt
sprays can significantly cause corrosion. The standard practice is to use piles which have a
factory applied epoxy coating. The most vulnerable sections of the piles should be encased in
concrete.
Hard driving and driving through obstructions causes the piles to twist and bend.
They can easily go out of plumb without the piling team recognizing since the depth is at
depth. Deviations from the vertical of below 10% are usually accepted. A penetration of 2 to
2.5mm per blow should be considered as refusal and further driving would generally cause
deterioration.
Pre-cast Concrete Piles
Pre-cast Concrete Piles (Figure 2.2b) are usually cast in a casting yard and transported to the
construction site. Where hard driving is expected the tip of the pile is fitted with a driving
shoe. They are usually of square or octagonal section. The reinforcement is necessary
within the pile to withstand both handling and driving stresses. It is necessary that the exact
length to be installed be determined accurately. If the required length is underestimated, the
extension can be done only with a lot of difficulties. If the length provided proves to be
longer than needed at the site, the piles have to be cut again with a lot of difficulties.
Pre-stressed concrete piles are used and generally have less reinforcement. The prestressing reduces the incidence of tension cracking during handling and driving. The
difficulties related to the pre-cast concrete piles also apply to the pre-stressed concrete piles
Pre-cast concrete piles have relatively large bearing capacity of between 800 and
2,000 kN per pile. The presence of high concentrations of magnesium or sodium sulphate in
the piled environments causes the piles to deteriorate. The deterioration is in the form of rust
in the reinforcement, cracking and spalling. The best practice is dense concrete of high
quality or the use of pre-stressed piles which are not so much susceptible because tension
cracks are minimized.
Driven cast in place piles
Driven cast in place fall in two categories namely case or uncased type. In the cased type
also known as shell the shell type a corrugated steel or pipe which is driven into the ground.
The driving is terminated when the desired length of the pile has been achieved. The
concrete is poured in the shell and left place. In the shell is then left in place. Figure 2.3
shows the schematic installation of a shell type pile.
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Deep Foundations
(1) RC
shells
threaded
on
mandrel and set in position
(2) Pile driven to the required
set
(3) Mandrel is withdrawn and
top shells above the top of
(1)
the pile are removed. A
cage of reinforcement is
introduced
(2)
`
(3)
(4)
(4) Core concrete is inserted
Figure 2.3 Shell type of pile
In the uncased type a steel tube is driven into the ground and tube is withdrawn upon
concreting. Figure 2.4 shows the schematic installation of a typical driven cast in situ pile
where the casing is withdrawn. The pile illustrated is also known as a Franki pile.
(1)
A gravel pug is compacted at
the lower end of the pile tube
(2) Pile driven to the required
set
(1)
(3) Plug broken and a concrete
plug is formed
(4) Core concrete is inserted
(2)
(3)
(4)
(5)
(5) Tube
is
withdrawn
as
concrete is placed
Figure 2.4 Installation of a Franki pile
Difficulties encountered in the installation of driven piles
The installation of driven piles has difficulties due to various factors incidental to the
installation procedures and to the ground encountered at the sites. These difficulties are
varied but the main ones include:a)
Handling of the preformed sections which could lead to damage of the piles before
installation.
University of Nairobi –FCE 511 Geotechnical Engineering IV
- 51b)
c)
d)
Deep Foundations
Noise arising from the hammer dropping on to the pile. This can be particularly
undesirable in sites in the busy neighborhoods.
Spoiling of the pile in the driving operations include the spoiling of pile heads and or
pile toes. This usually takes place due to overdriving piles when refusal has been
reached. It is usually sufficient to achieve a penetration of 2-2.5 mm per blow in the
last stages of piling.
Piles of small cross-section especially H piles driven in boulderly strata could easily
alignment. Vertical piles could end up having bent up shapes and hence lose their
carrying capacity.
2.1.4 Bored piles
Bored piles are also known as cast in place concrete piles (Figures 2.2c-e. The borehole is
effected by various methods using piling equipment. The bore is supported by casing or by
drilling mud (bentonite suspension). At the required depth boring is stopped and the hole is
filled with concrete. If required a cage of reinforcement is placed before concreting is done.
With the use of bored piles larger diameter piles have been installed with corresponding high
bearing capacities. They are constructed in diameters ranging from 300mm to as high as
2400mm.
They have been performed to depths of 70 metres and below and can be
constructed vertically or in rakes of up to 1:4. They are thus ideal for many site conditions.
The construction sequence of bored piles depends on the method of construction adopted.
The main construction methods include bored piles with casing support and bored piles with
bentonite support.
Bored piles with casing support
In this type of pile the casing is advanced by a crane and a casing oscillator. The material
below the casing area is excavated and brought up for examination and testing where
necessary. After the depth needed has been achieved the reinforcement cage is inserted
followed by concreting as shown on Figure 2.5
Bored piles with bentonite support
In this type of pile a lead casing is advanced into the soil. The material below the casing area
is excavated and brought up by use of drilling equipment with a bucket which can bail out the
drilled soil. The excavated soil is examined and tested where possible. The drilled hole is
supported by drilling mud After the depth needed has been achieved the reinforcement cage
is inserted
Figure
2.5 followed by concreting as shown on
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Deep Foundations
Install casing
Advance the
Insert
Place concrete with a
Complete
using an
casing and
reinforcement
tremie pipe as casing
pile
oscillator
excavate with grab
cage
is withdrawn
This installation is particularly desirable in gravelly and boulderly conditions
a)
With casing
Install
starter
casing
This
b)
Advance into the
Insert
soil by drilling
reinforcement
and supporting
cage
installation
is
suitable
in
all soils
with bentonite
Place concrete
with a tremie
pipe and recycle
bentonite
Complete pile
With betonite support
Figure 2.5 Installation of a bored pile with drilling mud
Difficulties encountered in the installation of bored piles
The difficulties associated with the installation of bored piles are also varied but the main
ones include:i.
Poor base preparation after the bearing strata has been reached. Loose particles will
have reached the bottom of the bore and will be difficult to detect or remove. The
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ii.
iii.
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base the pile will consequently have a lower bearing capacity than would have been
expected
Poor concreting control where the pile is being cast under artesian conditions. This
usually results from poor shaft control as the concreting continues. The result is
necking of the concrete and/or washout of various sections of the pile. Under ideal
conditions the concreter under tremie conditions should always be placed inside the
wet concrete.
Vibration and movement of the ground in the vicinity of the pile under construction.
It is to be noted that these difficulties are also present in the driven cast in place piles where
the casing is withdrawn as concreting proceeds
2.1.5 Determination of pile load carrying capacity
Determination of load carrying capacity by soil mechanics
Pile design is preceded by extensive site investigation to establish the geotechnical properties
of the soil where the piles will be installed. The parameters obtained in the investigations are
then used in the estimation of the load carrying capacity of the piles. Piles derive their
capacity from base resistance and from side friction. The ultimate load that can be carried by
a pile is then given by Equation 2.1. The terms are explained in Figure 2.6. The accuracy
of the equation depends on the determination of the parameters used in the determination of
Qb and Qs.
Where
= Ultimate Load carrying capacity of the pile
Ultimate Load carrying capacity of the base of the pile
= Ultimate Load carrying capacity of the pile side friction
2. 1
Where
Ab= Area of the pile at the toe of the pile
qf = Ultimate bearing capacity at the toe of the pile
= Surface area of the pile shaft
= Ultimate shearing resistance of the shaft of the pile generally referred to as the
shaft friction
An appropriate factor of safety is applied to the ultimate load. It is prudent to apply different
values for the base and the side friction. This is primarily because the movement needed to
mobilize the friction resistance is much less than the movement needed to mobilize the base
resistance. Initially as the pile is loaded the load is taken by the side friction and as load is
increased the base takes more load. At failure the proportion of load supported by friction
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may actually decrease slightly due to plastic flow of the soil near the base of the pile.
Equation 2.2 shows the allowable load when allowing for a factor of safety of 2 and 3 for side
friction and base resistance respectively.
2.2
Qs
Qb
Figure 2.6 Load distribution of load on a pile
Cohesive soils
Base resistance: The base resistance Qb of piles in cohesive soils is based on the bearing
capacity factor Nc .
2. 3
Where
= bearing capacity factor which is usually taken as 9.0
= undisturbed un-drained shear strength of the soil at the base of the pile
= the cross section area of the pile at the base
In the case of driven piles the clay adjacent to the pile is displaced both laterally and
vertically. Upward movement of the clay results in heave of the ground around the pile and
can cause reduction of the bearing capacity of the pile. The clay in the vicinity of the pile is
completely remolded during driving. Excess pore water pressures are set up during driving.
This pore pressure dissipates in a few months and in any case before significant load is
applied to the pile
In the case of bored pile, the clay area around the pile will be remolded. Additionally
as the water seeps towards the created borehole their softening of the soil in the vicinity of
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the pile. Water can also be absolved from the wet concrete when it comes in contact with the
clay. The upshot of this is and subsequent reduction of the pile bearing capacity.
Side resistance is based on the friction mobilized on the surface of the pile. Equation 2.4 and
2.5 shows the estimation of the side friction
̅̅̅
2. 4
̅̅̅
2. 5
Where
= adhesion factor between the pile and the soil
̅̅̅ = the average undisturbed shear strength of soil adjoining the pile
= the shaft area which contributes to the friction resistance
Most of the load of a pile installed in a clay soil is derived from the shaft friction and the
problem usually revolves accurate determination of the value of α. For soft clays driving of
piles tend to increase strength around the pile. A value of α equal to 1 can be used. It is
however unlikely that the soil will not in the long run return to its original soft status after
some time. In over-consolidated clays the value varies from 0.3 to 0.6 (Smith and Smith,
1998). A value of 0.45 is usually used for design purposes.
An alternative is approach is to express skin friction in terms of effective stress. The
rationale of this approach is that the area of disturbance during pile installation is relatively
small. The excess pore water pressure induced in the installation process dissipates ahead of
the application of load.
̅̅̅
́
2. 6
Where
Ks = the average coefficient of earth pressure and
̅̅̅ = the average effective overburden pressure adjacent to the pile shaft
́ = the angle of internal friction of the remolded clay. The cohesion intercept of
remolded clay in an drained triaxial test being zero.
Cohesionless soils
Base resistance: The ultimate bearing load carried by a pile depends mainly on the relative
density of the sand in which it is driven. The ultimate bearing capacity at the base of the pile
is given by
̀
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Where
= The bearing capacity coefficient.
̀ = The effective overburden pressure at the base of the pile
Value of Nq
It is to be noted that the bearing capacity attributable to Nγ usually ignored in pile design as
the value of B is usually small. The values suggested by Berezantzv et al (1961) are often
used and are shown on Figure
100
Nq
10
25
35
45
φ in Degrees
Figure 2.7 Bearing capacity factors for use in pile design
Source Berezantzv et al 1961
Side friction: Meyerhof (1959) suggested the average value of friction to be estimated from
Equation 2.6. As can be seen from the Equation the value of fs continues to increase as the
effective overburden increase. However field tests have shown that the maximum value of fs
occurs when the embedded length of the pile is between ten and twenty diameters. In practice
a maximum value of 100 kN/m2 of fs is taken.
̅̅̅
2. 7
Where Ks = the average coefficient of earth pressure and
̅̅̅ = the average effective overburden pressure adjacent to the pile shaft
= the angle of internal friction between the soil and the pile.
Typical values of and Ks are given on Table 2.1 after Smith and Smith (1998) are shown on
Table 2.1. The ultimate load that can be carried by the pile is therefore given by Equation
2.7.
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Table 2.1 Typical values of
and Ks
Pile material
Ks
Loose
0.5
1.0
1.5
o
Steel
Concrete
Wood
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20
0.75υ
0.67 υ
Dense
1
2.0
4.0
Source Smith and Smith (1998)
̀
̅̅̅
2.8
Equation 2.8 shows the allowable load when allowing for a factor of safety of 2 and 3 for
side friction and base resistance respectively.
̀
̅̅̅̅
2.9
Determination of piling parameters from in-situ tests
The above equations pose difficulties with respect to determination of parameters for a
cohesionless soil which is difficult to sample in the field in undisturbed condition for accurate
determination of Nq which depends on the internal angle of friction. The value of the angle of
internal friction between the soil and the pile remains at best an estimate.
Consequently it has been found preferable to use empirical correlations based on the
results of standard penetration and those of the Dutch cone penetration equipment. Meyerhof
(1976) proposed the values given on Table below.
Table 2.2 Pilling parameters from standard penetration tests
Driven piles
Type of soil
Sands and gravels
qb (kN/m2)
Non plastic silts
Bored piles
Any types of soils
fs (kN/m2)
Large diameter - ̅
Average diameter - ̅
Large diameter - ̅
Average diameter - ̅
0.67 ̅
Source Smith and Smith (1998)
Where N = the uncorrected blow count at the base of the pile
̅ = the average uncorrected value of the blows over the embedded length of
the pile
D = is the embedded length of the pile in the bearing stratum
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B = the width or the diameter of the pile.
An alternative to the use of the Standard Penetration tests is to use the Dutch cone test results.
The cone penetration results can be seen in
Figure 2.8. The ultimate base resistance is taken as average value of Cr over a depth of 4d as
shown on
Figure 2.8. The ultimate skin friction can be obtained from Table 2.3.
3d
of the pile
Estimated depth
Depth (m)
Cr (kN/m2)
d
Figure 2.8 Typical results from a Dutch Cone Test
Table 2.3 Skin friction (fs) values from Dutch cone test results
Type of pile
fs kN/m2
̅̅̅
Driven piles in dense sand
Driven piles in loose sand
̅̅̅
Driven piles in non plastic silts
̅̅̅
Where
̅̅̅ is the cone resistance along the embedded length of the pile
The allowable bearing load of the pile as before based on the Dutch Cone Test results is
given by Equation 2.9
2.10
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2.1.6 Determination of load carrying capacity dynamic methods
Determination of load carrying capacity dynamic methods is applicable to driven piles. The
basis of derivation of dynamic formula is that a relationship exists between the pile capacity
and the driving behavior during the last stages of driving. The energy from the hammer to the
pile is transformed into useful energy and can be represented by Equation 2.10 in the last
stages of the pile driving
2. 11
Where
M = the mass of the hammer
g = the acceleration of the hammer
h = the drop the hammer
R = the pile capacity
S = the settlement of the hammer as result of the drop h
In practice the above Equation has been modified to take account of several losses which take
place during the driving process. The main losses of energy occur as a result of sound, heat,
friction, quake, losses associated with elastic behavior of the pile and those associated with
the pile head compression. The net energy is equated to the work done in penetrating the
ground by the pile. Figure 2.9 shows the sequence of the pile driving and the
Wh
h
e fW
efeivWh
h
a) Variation of energy upon falling of hammer on to a driven pile
Permanent +Elastic penetration
(sso + spp) +(sep +ses)
(sso + spp =set =s)
(sep +ses )=c)
(sso+ses)
(sso)
(ses)
b) Penetration of pile upon falling of hammer on to a driven pile
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Figure 2.9 Energy and penetration of a pile during driving
The potential energy of the hammer is Wh. Upon contact with the pile the available energy to
drive the pile into the ground is ef.eiv.Wh, where ef is the efficiency upon falling and eiv is
the efficiency upon impact. The penetration of the pile as shown on Figure 2.9b can be
shown to result in permanent ;penetration attributable to the pile and soil spp aand sso. In
addition there will be elastic penetration sep and ses attributable to the pile and soil
respectively. The work done and the pile resistance equation can now be rewritten as shown
on Equation 2.11.
2. 12
Where R = The ultimate load capacity of the pile
= the overall efficiency factor
Equation 2.10 is known as Hiley formula. In the field the final stages of the pile are
monitored and recorded as can be seen on
Figure 2.10. It is usual to drive the piles to a minimum set of 2.5mm. Harder driving only
goes to damage the toe of the pile and could reduce the pile capacity in the process. Pile
driving formulas should be used in the piles driven in sand and gravel and in any case should
be calibrated with a load test.
Elastic comp = c3
Elastic comp = c2
Elastic comp = c1
set = s1
Figure 2.10 Pile driving trace of the final stages
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set = s2
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2.1.6 Determination of load carrying capacity pile testing
The load test is the most reliable of all the methods used in the determination of load carrying
capacity of a pile. In this method a full scale test is carried out on a working pile. Essentially
the pile is loaded and a plot of load versus settlement is recorded. From the plot the
allowable load is computed by one of the many formulas available from literature. Full scale
piles are then constructed to the same specification as the test pile
The test is conducted by loading the pile with kentledge load or by use of tension piles
(Figure 2.11). In some piling contracts the working piles cannot be used as tension piles for
testing purposes. This is primarily because in the cause of piling test the tension piles are
lifted slightly. This could lead to weakening of the working piles.
Kenteledge
Kentledge
Support
Existing ground level
Jack
Test pile
a)
Load resisted by kentledge
Jack
Existing ground level
Test pile
Tension pile
b)
Tension pile
Load resisted by tension piles
Figure 2.11 Methods of testing piles in the field
If the test pile is a purely test pile ahead of the main installation of the pile the maximum load
to be applied is equal to two and half times the estimated safe carrying capacity of the pile.
It is usual to load the pile to 1.5 times the design allowable pile load when a working pile is
tested for ascertaining the integrity of the piles installed.
Maintained load test
The load is applied by maintaining the load in a series of increments. The increments are
usually equal to 20 to 25percent of the design working load of the pile. The subsequent
increments are carried out when the settlement has reduced to less than 0.25mm per hour.
The load is subsequently withdrawn in the same stages as the loading to trace the unloading
curve.
Constant rate of penetration
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Load
In this method the load is applied by a constant rate of penetration by a jack in order to
maintain a constant penetration rate (Figure 2.11b). it is usual to maintain penetration rates
of 1.5mm per minute and 0.75mm per minute in the case of sands and clays respectively.
Interpretation of test results
The results are plotted on a load settlement curve as shown on Figure 2.12. In the two
procedures ultimate pile load is taken as the load which achieves a settlement equal to 10
percent the diameter of the pile as is seen in test pile a Figure 2.11b. (BS 8004). The ultimate
pile load could also be reached when the shear failure of the pile soil interface or the pile toe
occurs (Figure 2.12b). The allowable pile load is obtained by dividing the ultimate load by
an appropriate factor of safety. The factor of safety usually ranges from 1.3 to 2.0
Load
Settlement
Settlement
Time
a) Maintained load test results
a
Ultimate
load (a)
Load
Ultimate
load (b)
b
Penetration =
0.1 pile diameter
Penetration
b) Constant rate penetration test results
Figure 2.12 Pile test load results
The above failure criterion is applicable to normal size piles. In the case of large diameter
piles on rock the ultimate load depends on the capacity of the concrete. This depends on the
stress in the concrete.
2.1.7 Negative skin friction
Negative skin friction is a phenomenon or which occurs in piles when a force develops
between the pile and the adjoining soil in a direction which increases the load on the pile and
or the pile groups. This phenomenon develops when a compressible layer of clay, silt, or
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mud etc settles on account of consolidation which may be initiated by ground water lowering
or increase in overburden pressure.
As clay layer settles, piles are dragged into the soil by the consolidating soil and the
overburden soil. The direction of the friction is reversed increases the load on the pile. The
friction generated on the perimeter of the pile due to this dragging is carried by the column
instead of assisting in carrying he pile load. The effect is to reduce the carrying capacity of
the pile. This is the phenomenon known as negative skin friction
l-fill
Fill
Compressible clay
l-clay
Length of settling soil=l
Figure 2.13. The negative skin friction may be estimated from Equation 212 for
single piles and Equation2. For group piles
Figure 2.13 Negative skin friction
2.13
For cohesive soils fs is can be approximated to
̅ . while for cohesionless soils fs is
equal to
̅ . Where the value of fs is estimated from triaxial testing for cohesive
soils the fs can be taken as 0.5Cu
Where
= the ultimate force generated by the negative friction
= the shearing resistance of the soil
= length embedded above the bottom of the compressible layer
= the pile diameter
= the coefficient of earth pressure at rest
= angle of shearing resistance in terms of effective stress
̅ = average effective overburden pressure
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2.1.8 Pile groups
In practice piles are designed and constructed to work in groups. In construction of a group a
pile cap is cat on top of the piles. The cap is usually in contact with the soil on top of the
piles. The bearing capacity of the group is an arithmetic sum of the piles and that of the cap.
Banerjee (1975) showed that the pile cap could support up to 60% of the applied load. If the
cap is clear of the ground surface piles in the group are referred to as free standing piles.
Bearing capacity of groups
Except for the large diameter piles of over 700mm diameter the piles are usually designed in
groups of three or more piles under a column. The minimum under a foundation wall would
be two per typical cross-section. Typical arrangement of the piles is given on Figure 2.14. In
general the ultimate load capacity of the pile group is not the sum of the loads of the piles in
the group. The ration of the ultimate load for the group to the sum of the loads carried by
individual piles is the efficiency factor of the group.
3 – Pile
4 – Pile
5 – Pile
12 – Pile
Figure 2.14 Typical arrangement of pile groups
For piles in sand, the group action is complicated by dilatancy and densification
characteristics of the sand. When the spacing of the piles is less than eight times the pile
diameter, group action takes place (Department of Navy, Naval Facilities Engineering
Command, 1982). In dense sand the effect of driving piles is to loosen the sand and hence
the angle of internal friction of the sand in the vicinity of the piles. This results in overall
reduction of the pile bearing capacity. The group efficiency factor is less than one. In loose
sand the effect of driving piles is to increase the density of the sand. The bearing capacity of
the loose sand will therefore be increased. In this case the efficiency factor is more than one.
An efficiency factor of 1.2 is often used. In the case of bored piles in sand the resulting
loosening of sand in the boring operation results in efficiency factors less than 2/3. The
difficulties in the quantification of the design parameters of either loosened or densified sand
strata in piling operations remains a real problem for engineers (Mwea, 1984). Nonetheless
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experimental evidence has it that the piles at the centre of a group in sand carry more load
than the piles on the periphery.
For piles in clay the effect of the pile group is to reduce the bearing capacity of the
pile group. This is because the effect of placing piles in a group is to have one large block
taking friction on the sides and base resistance over the block base. The spacing of piles in
clay is of the order of two times the pile diameter to four times the diameter. The efficiency
of the groups range from 0.6 to unity as the pile spacing increases from two diameters to four
diameters. The ultimate load in the case of a pile group is given by Equation 2.13. In the
case where the pile cap rests on the ground the ultimate load should be taken as the less of the
block capacity or the sum of the individual piles on the group.
2. 14
Where
= The width of the group
= Length of the base of the group
= Depth of the group
= Bearing capacity factor of the clay
= The average undrained strength of the undisturbed clay
Whitker (1957) in a series of model tests showed that block failure as a group in clays occurs
when the spacing of the piles is not more than 1.5d apart. General practice is however to
space the piles at between 2 and 3d. In such cases the efficiency of the group is
approximately 0.7.
Settlement of groups
The settlement o a group of piles can be estimated by assuming that the entire load acts at a
depth as an equivalent raft. In clays the raft is assumed to be located at a depth of 2/3 D
where D is the depth of the pile group. The load is at spread of 1:4 from the underside of the
pile cap to allow for friction transfer. After the assumed depth of the raft the load is
distributed at a spread of 1:2 (Error! Reference source not found.a). Immediate settlement
and consolidation settlement can then be estimated for the layers of soil below 2/3D by
application of normal methods.
For groups in sand the equivalent raft is at a depth of 2/3Db from depth 2/3D. The
spread from the perimeter of the piles is 1:4 followed by a spread of 1:2 Error! Reference
source not found.b). The settlement of the underlying sand stratum is then gotten from
application of standard penetration data and or the cone penetration resistance
1:4
D
1:2
2/3D
2/3D
Db 1:4
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1:2
Position of equivalent raft
Position of equivalent raft
Clay stratum
Sand stratum
Figure 2.15 Equivalent raft concept for piles
2.2
Drilled piers and Caisson Foundations
2.2.1 Drilled piers
The term drilled pier foundations is used in a number of situations which to refer to deep
foundations which method of construction is fundamentally different from that of piles. A
large shaft performed in soil and then filled with concrete may be termed as a drilled pier.
ACI (1972) refers to all shafts where a person may enter and work as a drilled piers. In this
definition all shafts larger than 750mm diameter can be referred to as drilled piers. Figure
*** shows typical piers used in practice. In general drilled piers are used where the soil has
a low bearing capacity and it is necessary large loads to firmer stratum and the following
conditions preclude the use of smaller piles.
i.
ii.
iii.
Pile vibrations are not acceptable.
Pile members are too small for the loads.
A large bearing end is needed for higher load capacity
Straight pier
Underreamed pier
Pier socketed Into
Rock
2.2.2 Caisson Foundations
The term caisson is also used to refer to box type structures consisting of many cells built in,
concrete or steel or combination of both. They are built wholly or partly at higher ground and
sunk to final position. They are used to transmit large loads through water and soil to firm
strata. They are used in large bridges, shore protection structures. They are generally used
under the following conditions.
i.
The soil contains large boulders which would otherwise obstruct the penetration of
piles and or construction of cast in place piles.
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A massive substructure is needed to extend below the river bend to provide resistance
against floating objects and scour.
Foundation is subjected to very large lateral forces.
iii.
Caissons may be divided into three categories
i. Open caissons
ii. Pneumatic caissons
iii. Box caissons or floating caissons
Open caissons
An open caisson essentially consists of a box open at the top and bottom ( Figure 2.16). the
soil is removed from the caisson by grabbing, dredging from inside the caisson. The sinking
of the caisson proceeds by the caissons self weight assisted by cutting edges of the walls.
When the desired level has been reached concrete is poured under onto the base of the
caisson by tremie pipe. In some cases the caisson has been pumped out. But in most of the
cases the caisson has been left in place. The bearing capacity of the soil below is usually
determined by normal bearing equations.
The concrete seal at the bottom is placed as a plug at the bottom of the caisson but
later serves as a permanent base of the caisson. Its thickness can be obtained from the
equations below
For circular caissons
√
For rectangular caissons
√
Where
= thickness of the seal
σo = contact pressure or hydrostatic pressure
R = radius of the caisson in the case of circular caisson
fc = the allowable concrete stress in tension (0.1 to 0.2cube strength)
b= width or the short side of the caisson in the case of a rectangular caisson
l= length or the long side of the caisson in the case of a rectangular caisson
β = coefficient which depends on the l/b ratio
Water level
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Ground surface
Cutting edge
Box caisson
Circular open caisson
Figure 2.16 Open Caissons
Pneumatic caissons
Pneumatic caissons provide an airtight enclosure (Figure 2.17). In effect water is prevented
from getting into the enclosure and the workers can excavate and pour concrete under dry
conditions. The reliability of the quality in this case is better in so the mechanical ventilation
is carried out to the strictest of the specifications. Pneumatic caissons are costly and should
be considered only with the following conditions in mind:
i.
ii.
iii.
Premium pay because of associated health hazards
Overall safety requirements are high
Much of the effort is towards making the work environment suitable for the workers
When the excavation has reached the desired stratum the concrete is sent down to the
working chamber carefully to fill any weak points on the exposed strata. After this initial
filling the area is filled except a small portion of the chamber below the roof of the chamber.
This final portion is filled with grout which also fills any spaces which might have been left
behind during the concreting.
The seal design and estimation of the bearing capacity is the same as that of the open
caissons
Compressed air in
working chamber
Figure 2.17 Pneumatic caissons
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Box caissons
Open caissons are usually cast on the ground and then towed to the site. They area then
lowered to a prepared ground. They are carefully aligned on place and then made stable by
placement of ballast. The design and construction of box caissons do not bring any new
design requirements. The ground upon which the caisson is being laid needs to have been
exhaustively investigated to ascertain the foundation depth and any likely difficulties likely to
be encountered. After the caisson is in place it may be filled with either sand concrete or
sand. The caisson should be checked against stability as it is floated to the final place of the
intended foundation.
Design of caissons
The caissons will be designed to resist vertical loads including superstructures, own weight
minus buoyancy forces. The lateral forces will typically include forces due to wind,
earthquake, earth and water pressures, and traction from traffic and pressure from current
flow.
The forces acting on a caisson must be estimated as accurately as can be to enable a
safe design. There are many methods adopted by various geotechnical engineers but the for
stability of the caisson the following combination of forces will suffice
i.
ii.
iii.
All forces are resolved into
A single vertical force
Two horizontal forces in the direction across and along the caisson.
It has been found out that analysis of the caisson in a direction transverse to the direction of
the axis is more critical. From Figure ***-* the three equations of static equilibrium are
solved. This are
W = Base reaction + skin friction
Q = Passive pressure created on BF – Passive pressure on DE – Base friction
Q (H+D) = Moment of all the forces
Q
Q
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From structural
analyses
h
D
From geotechnical
analyses
O
From structural
W
analysis
Q
From
geotechnical
analysis
H
A
D
F
D1
O
E
B
γD(Kp-Ka)
C
γD(Kp-Ka)
Qmax =Area ABC-Area FEC
Qmax =1/2 γD2 (Kp - Ka)- ½*2* D (Kp - Ka)*D1
Moments about O:
Qmax (H+D)=1/2 γD2 (Kp-Ka)D*1/3- ½*2* D (Kp-Ka)*D*D1*1/3
Therefore D1 and Qmax can be calculated and necessary adjustments of the caisson are
made depending on values of Kp and Ka
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2.4
Deep Foundations
Examples of Piling Schemes
Sutong bridge in China
Sutong bridge in China (Plate 1), which has a centre span of 1088m, designed in an area of
high winds and likely to be hit with massive earthquakes (Bitener et al, 2007). The
foundation strata presented the designers with particularly difficult task. The soils at the site
consisted of firm to stiff clay extending to 45 metres below the sea bend. This clay strata was
underlain with a medium to very dense coarse sands, silty sands and occasional loam layers
matrix to a to of 250 metres below the sea bed where the basement rock was encountered.
The designed pile groups covered a plan area of 113.8x48.1m. The design consisted
of 2.8 and 2.5 diameter piles. Permanent casings were installed to a depth of 40 metres. The
overall depth of the piles was of the region of 110 metres. The shafts were designed to mainly
be carried by friction since the displacement needed to mobilize the end bearing is two to
three times that needed to mobilize the skin friction The tips of the pile shafts were however
grouted to increase the bearing capacity of the piles. This procedure densifies the soil below
the shaft and any debris left during the drilling operations. The increased the pile capacity
end bearing capacity is of the order of 20%.
Plate 1 of the Sutong Bridge in China (1088 m center span)
The Nyali bridge in Mombasa
This is a pre-stressed concrete bridge founded on seabed which had coral deposits, sand and
clay soils matrix proved to a depth of 100metres below the sea bend. The designers
depended on the skin friction for the centre piers. The design consisted of 2.0metre diameter
shafts drilled down to depth of 50 metres. On plan the piles have a rectangular layout of 3x8
piles per pier.
2.5
Tutorial examples on chapter two
1)
A single pile 0.6 m diameter is bored into sand strata six meters thick overlying a clay
stratum of infinite depth. Detailed investigations have established that N value in the
sand zone increases with depth (n=3Z). The undrained cohesion increases with depth
(Cu = 5+4Z). Assuming the adhesion factor α = 0.35, determine
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Deep Foundations
a)
2)
An equation for the estimation of pile working load if the pile is to terminate
in the sand zone.
b)
An equation for the estimation of the pile working load if the pile is to
terminate in the clay zone.
A precast reinforced concrete pile measured 450mm x450mm. The pile was driven to
a depth of 15 metres to a set of 3mm by a drop hammer of 2.5 tones freely through 1.5
metres. The piling arrangement was changed to have a 4.2 tone hammer falling
through 2 metres. Assuming the same resistance with the new hammer, determine the
set achieved if the following information is also available.
2.5 tone hammer
0.5
4mm
4.5mm
Overall efficiency factor
Elastic compression of pile
Elastic compression of soil
4.2 tone hammer
0.35
4mm
5.0mm
3)
A pile under test has started showing considerable settlement under load of seventy
tones. The pile diameter is 500mm and a length of 8.5metres in stiff clay. Assuming
below the 8.5metres the clay was soft clay and did not contribute to any resistance
evaluate the magnitude of the unit shear along its skin. (Answer 10.5tones per m2).
4)
A 500mm diameter bored pile is to be made in stiff clay to a depth of 20metres. The
un-drained strength of the clay varies with depth as shown in the following table
Depth
4
2
Cu (kN/m ) 78
6
86
8
102
142
132
16
157
20
184
24
212
Determine the maximum load that may be applied to the pile. The following factors
may be taken.
Adhesion factor α = 0.45
Overall factor of safety = 2
Nc for piles is usually taken as = 9
(Answer 1025kN).
University of Nairobi –FCE 511 Geotechnical Engineering IV
Chapter Three: Introduction to Earth Dams
3.1
Introduction
Advances in geotechnical engineering have enabled design and construction of high dams
impounding large amounts of water. The design and construction follows well documented
procedures gained over the last years from design construction and monitoring of both
successful and unsuccessful projects. The procedures now taken include
i)
Thorough pre-design and preconstruction investigation of the dam foundation
conditions and of the construction materials and design of dams.
ii) Application of engineering skills and techniques to design
iii) Carefully planned and controlled construction
iv) Carefully designed and installed instrumentation and monitoring of the completed
dams
The design and construction of a dam is not complete without accomplishing its intended
purpose and has proved it safe over several cycles of the performance. Carefully designed
and constructional dams are in excess of three hundred meters high. Our own Thika dam
which supplies the Nairobi residents with water rises some sixty three meters above its
foundation.
Failures in dams have been occasioned by improper design, inappropriate
construction methods, including preparation of foundations, placement of the dam
embankment layers, without the necessary controls of compaction control and monitoring.
The design and construction should not be stereotyped on existing dams. Rather each dam
should be unique and dependent on the geology of the available materials. As one embarks
on the design of dams it should the course of dams the causes of failure of dams has been
listed by Singh and Prakash (1985) as shown on Table 3.1
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Introduction to dam design
Table 3.1 Causes earth dam failures
Cause of Failure
Overtopping
Seepage effects (piping, sloughing etc)
Slope slides
Conduit leakage
Damage to slope protection
Miscellaneous
Unknown causes
% Occurrence
30
25
15
13
5
7
5
Type of failure
Hydrological
Geotechnical
Geotechnical
Geotechnical/Structural
Maintenance
General
General
Source: Singh and Prakash (1985)
The design and construction techniques covered in this chapter are applicable to all dams.
However the design and construction of small dams in Kenya is well covered in the manual
prepared by Ministry of Water (1985). Small dams are those whose height does not exceed
15 metres and or its impounded volume does not exceed one million cubic meters (Bureau of
reclamation 1985). The procedures covered in this chapter are inappropriate for the design
and construction of dam materials presenting the followings characteristics
i) Extremely soft, or dispersive or materials with high plasticity
ii) Exceedingly pervious foundations
iii) Exceedingly fractured foundations
These conditions require specialized testing and analysis of the presenting conditions in order
to arrive at an appropriate design
3.2
Selection of type of earth dam
The scope of dams covered is those dams where the major portion of the embankment is
constructed in successful layers compacted in layers. The layers are well bonded into one
another to achieve the necessary requirements of the particular layer. The materials are
borrowed from borrow pits and from the reservoir area of the dam. Earth dams fall into three
categories namely, diaphragm, homogenous and zoned
3.2.1 Diaphragm types
This type of dam is constructed with pervious materials namely sands, gravels and or rock.
An impervious diaphragm is constructed to act as the main barrier to seepage. The
diaphragm is usually made of concrete, or bitumen. Alternatively they are made of thin
compacted earth. In this case the width of the diaphragm at any depth is either less than three
meters or it thickness at any elevation is less than the height above that elevation. Figure ***
shows typical diaphragm type dams
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Introduction to dam design
As with all dams the diaphragm dam should be designed and constructed with care
and precision. All internal diaphragms whether made of rigid materials like concrete or even
compacted earth have potential of cracking caused by differential movement of induced
during consolidation of the dam embankment materials, fluctuating water levels or settling
foundations. Internal concrete diaphragms can not be readily inspected. Earth diaphragms
on the surface require protection with filters, protection against erosion and wave action.
These types of diaphragms are unusually protected by rock fill and rock riprap. The earth
diaphragm is also not readily inspected during routine or emergency inspections. The earth
diaphragms are usually protected from internal erosion by filters usually in the form of
geotextiles.
If most of the material in a diaphragm dam is rock, then this type of dam is referred to
as a rock dam discussed below.
3.2.2 Homogenous types
These types of dams are made up of single kind of material save for the slope protection. The
material in this type of dam must be sufficiently impervious to act as the barrier for the
seepage. Because the impervious materials are inevitably clays which are weak in stability
but good as barrier to the seepage the slopes tend to be rather flatfish. The usual slopes on
the upstream side of the dams are 1:3.5 to 4 while for the down slope slopes need slopes of
1:2.5 to 1:3. Figure *** shows a typical homogenous slope with three flow lines. As can be
seen, seepage inevitably appears on the downstream side at a height of about 1/3 of the height
of the dam.
Rock toes and horizontal blankets are usually used to avoid the seepage breaks on the
down slope side of the dam. Riprap protection is also used on the upstream side to arrest
erosion occasioned by the waves on the upstream side of the dam. Drainage and filter layers
are designed to meet filter requirements. Inclined filters in combination with horizontal filters
built with well graded sand and surrounded by geotextiles have become a normal practice.
Because modification of the homogenous dams has led to successful dams the use of
completely homogenous dams is now not allowed. The homogenous dams are preferred
where other materials of contrasting permeability are unavailable. Alternatively they should
be used where impervious material forming the embankment is abundant and available
principally in the dam area and within the vicinity of the dam.
Zoned types
In this dams, a central core is of impervious material is franked by more pervious materials.
The design of these dams requires that the permeability of dam embankment materials
increases from the core to the outside franking shells. The materials enclose support and
protect an internal impervious core. The upstream sections provide stability during rapid
drawdown. The downstream pervious materials act as drainage to control the line of seepage.
It is usual to place a filter material between the impervious material and the downstream
pervious materials.
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Introduction to dam design
The impervious inner layers are basically clays typically the red coffee soils. The
pervious layers are sands, gravels, cobbles, boulders and rocks. If a variety of soils are
available the type of dam of choice is the zoned dam (Bureau of reclamation 1985). It has
inherent advantages of stability and reduced seepage across the dam wall.
3.2
Design Principles
The dam should be constructed so that a satisfactory performance at minimum cost is
attained. The maintenance costs should also be factored to ensure a facility with the least
maintenance of the upstream, downstream and the apartment structures and the electro
mechanical structures. An earth dam must be stable during all phases of the construction and
the operation of the reservoir. To accomplish this, following criteria must be met:
i)
ii)
iii)
iv)
v)
vi)
vii)
The embankment, foundation, abutments, and reservoir sides must be stable and
should not develop unacceptable deformations during construction or during the usage
of the structure
Sufficient seepage control must be ensured to ensure that excessive piping, instability;
sloughing, material erosion is under control. Additionally the loss should be such that
it dose not impair the intended usage of the facility by excessive loss of water.
The reservoir sides should be stable under all operating conditions to prevent
landslides into the reservoir. It is to be noted that a landslide into the reservoir could
cause large wave to overtop the dam
The embankment must be provided with adequately sized spillway which allows
design flow floods to pass without overtopping the embankment.
Free board allowance should be sufficient to prevent waves from overtopping the
dam.
The dam should be provided with camber which allows settlement of the foundation
and the abutment to take place. This camber is not included in the freeboard
calculations.
The upstream slope must be protected against the wave action while the down slope is
protected against rain erosion and animal grazing
3.3.1 Foundation design
Foundations of dams refer to the dam embankment wall floor and the sides of the
embankment in touch with the original ground of the dam. Foundations are usually not
designed but they require attention to ensure satisfactory performance. The requirement of
the foundation is to be stable under all conditions and to offer sufficient resistance to seepage
to prevent loss of water.
To determine the seepage and stability conditions of foundations the permeability of
the foundations strata in various directions and at various depths need to be determined. In
addition the strength of the strata should be determined by use of appropriated field testing
accompanied by field testing. For small dams however it is normal to use empirical approach
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Introduction to dam design
in the treatment of the foundations. Because the foundations of different materials demand
different treatments the foundations are grouped into three different classes. These classes
can be grouped into
i) Rock foundations
ii) Foundations of coarse grained materials (sand and gravel)
iii) Foundations of fine grained materials (silt and clay)
Rock foundations
Ordinarily the rock foundations do not present any bearing capacity problems. Instead it is
the seepage problems which have to be addressed A thorough site investigation should be
undertaken to establish faults and any areas of excessive weathering which could lead to loss
of water. The procedure would be usually to perform in-situ tests to determine the
permeability of the rock structure. This is undertaken together with a site survey of the
fissures of the rock. If excessive erosive leakage , uplift pressures, high water pressures can
occur though rock crevices, fissures, permeable strata, and/or fault planes, consideration
should be made to grout the foundation.
The foundation grout is basically injection of a sealing material under pressure into
underlying formations through grout holes. Grout ordinarily consists of cement water
mixture in the ration of 10:1 in the case of rich mix to 0.8:1 in the case of a lean mix. Some
additives to the cement water mix is usually done to improve the pumping. The most used
additive is betonite
The injected grout eventually fills the cavities and potential avenues of water.
Grouting is a procedure requiring specialized personnel and equipment to effectively carry
out the operations. In general a centerline curtain of grout of holes spaced at three to six
meters is adequate. Where large zones of fracture occur below the dam wall and in the
immediate upstream of the dam a blanket grout on grid is desirable. The depth of the grouting
is usually in the region of three to ten metres. In most cases a blanket grouting of the
foundation directly below the impervious zone is desirable.
Sand gravel foundations
Generally these foundations have sufficient strength to adequately support the loads induced
by the embankment and the reservoir. Nonetheless exploratory and analysis of the strata
must be carried out as a matter of routine. The main problems of these foundations are under
seepage and subsequent forces exerted by this seepage. These undesirable effects should be
analyzed and mitigated in design and construction.
Foundations on looses sands are suspect and should generally be avoided as the sand
has the potential of collapse under load. These type of foundations should be avoided or
specialized advice sought.
The amount of under seepage should be estimated from values of coefficient of
permeability of the strata. The coefficient of permeability of the strata should be determined
by established methods including pump out tests, tests conducted by observation of boreholes
when pumping is performed in a test borehole or pump in tests as described in FCE 311. The
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Introduction to dam design
magnitude of the seepage forces should also be determined by analyzing the flow net of the
water flow under the dam. This topic has been covered in FCE 411.
The various methods of treatment of the foundations of sands and gravel should aim
at economical control of the under seepage and the control of the subsequent seepage forces
to prevent the undesirable effect of foundation erosion and piping at the exit of the dam.
Excessive treatment of a detention dam might not be necessary while treatment of foundation
of a water supply dam might be prudent. The various treatment techniques are now presented.
i) Cutoff Trenches
Where possible this is usually the treatment of choice. The cutoff should extend down to
bedrock or to other impervious strata. This treatment ensures no future difficulty will be
experienced in piping and or uplift of the dam. A minimum width of the cutoff trench is
shown on Figure 3.1
h
Sand gravel
d
w
Rock
Figure 3.1 cutoff trench
ii)
Partial cutoff
A cutoff that does not go all the way to the foundation, rather it is designed to a proportion of
the depth to the rock or to impermeable layer.. the reduction in area is not proportional to the
reduction in the flow. Thus the reduction cannot be estimated from the flow equation
The action of the partial cut off is similar to that of an obstruction of in a pipe. The reduction
in flow is not proportional to the reduction in the area of the pipe. Experiments have shown
that a 50% cutoff results in 25% reduction of the seepage while an 80% cutoff results in 50%
reduction in the seepage.
iii)
Sheet piling
This is an expensive method of cutting of the seepage through the foundation of an earth
dam. Additionally the seepage continues to pass through the sheet pilling interlocks. It has
been used sometimes in conjunction with the cutoff trenches. The sheet piles cannot be
performed in cobbles and boulders
iv)
Slurry trench
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Introduction to dam design
This is a trench excavated and filled with concrete below the impervious layer. The trench is
kept in position by placing bentonite before concreting to form
v)
Grouting
Various materials have been used to develop grouting procedures to improve the stability and
reduce the permeability of pervious foundations under dam walls. These materials include
a)
b)
c)
Cement – water
Cement – bentonite – water
Bentonite
It is to be noted that grouting is usually an expensive process and it should be allowed after
extensive testing and evaluation.
vi)
Upstream blankets
These are usually made of same material as the impervious core material. In effect the path
of the underseepage is increased and hence the loss of water is reduced.
vii)
Downstream embankment toes
The aim of these blankets
a)
To reduce uplift pressures at the exit of the dam
b)
To readily permit discharge
c)
To prevent piping of the fines
d)
To convey the discharge
Achieved by
a)
Extending the downstream zones
b)
Pervious downstream shells or by use of horizontal blankets
c)
By use of toe drains
3.3.2 Embankment Design
Embankment slopes
The design of and earthfill dam embankment needs combination of many parameters. Many
of these parameters are difficult to determine accurately. These parameters include gradation,
composition and corresponding behavior of the soils under different conditions of saturation
and loading. The stress – strain relationships can be very complex. The result of these
difficulties is that the design of earthfill dam embankment depends largely on successfully
designed, constructed and well performing dams.
Notwithstanding these acknowledged difficulties it is now possible to model out complex
conditions of an earthfill dam embankment. The design of any earthfill dam is preceded by
extensive site investigation to determine the strength and permeability characteristics of the
embankment materials. This enables the design of h the slopes to the embankments to be
checked under the follow conditions.
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Introduction to dam design
i)
Stability during construction and end of construction. In this condition the
embankment has not had the time to have the pore water in the foundations and the
embankments drained. The pore water pressures are highest in the embankment
materials. The strength parameters applicable are the undrained parameters.
ii)
Steady seepage conditions. The core of the dams act as the water barrier of the earth
fill dam. However even the tightest of the clay cores will allow some water
penetration. The rate of penetration will depend on the permeability of the core
material an in due time will reach steady seepage conditions where a phreatic surface
will be developed at the highest level in the embankment. The steady seepage
conditions is critical for the downstream slope. Under these conditions the water has
been impounded the seepage has stabilized through the embankment. The flow net
has been established. All the excess pore water pressures have dissipated. The slopes
of the dam are checked using drained parameters of the foundations and the
embankment materials. The downstream slope is in critical condition during the
steady seepage
iii)
Rapid drawdown conditions. Under these conditions the stabilizing effects of the
water in the reservoir has been removed on the upstream slope. The rapid drawdown
leaves high pore-water pressure in the embankment. The upstream slope is usually in
its weakest state. The upstream slope of the dam is checked using drained parameters
of the foundations and the embankment materials. It is to be noted that a drawdown
of up to 40 meters per day is considered as rapid.
iv)
Stability under severe seismic conditions. The above conditions should be subjected
to acceleration of the embankment occasioned by seismic activities
v)
Protection against erosion. The upstream slope is likely to be subjected to erosion
arising out of the wave action and sloughing as the level of the water fluctuates. This
is mitigated by use of appropriate upstream protection by use of stone riprap. The
downstream slope is subjected to erosion a result of the precipitation and made worse
by grazing in some dams in communities in need of pasture. The usual practice is to
fence off the dam area and to plant grass and appropriate trees.
The stability check is usually to ensure that the shear stresses induced in the embankments
are resisted by the mobilized shear strength. The shear stresses are from the externally
applied loads which include reservoir weight and earthquake forces. Additionally internally
generated forces from the self weight of embankment The shear stresses at the slopes being
checked can be shown on Figure **** below the shear stresses to be resisted is shown on
Equation 3.1
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Introduction to dam design
σ1
σ
τ
σ3
θ
̅
3.1
The external and internal forces produce a compressive stress along the sliding surface. This
mobilizes the shearing strength which resists shearing along the surface being checked. The
shearing strength is given by Equation 3.2
̅
3.2
It is to be noted that while the shear strength is reduced by the increase in the pore water
pressure the shear stress remains the same. This shows the need of understanding and taking
care of the changes in the pore water regime. In practice the design involves the checking of
the slope stability and application of a suitable factor of safety
Compaction
Compaction of earthworks is a key activity to ensure that the envisaged strength and water
tightness is achieved. When the compressibility and loading of the embankment are constant
the more saturated the soil is the higher the likely hood of developing high pore-water
pressures. To minimize the development of high pore-water pressures it is compact the
earthworks just dry of optimum. However for low dams it has been found satisfactory to
compact earthworks at MDD and OMC.. At this moisture content the material is able to
conform to the shape of the foundation and the abutments.
3.3
Inspection of existing dams
General appearance
i)
Sagging crest
ii)
Slope failures
iii)
Wet patches
iv)
Slope protection
v)
Soil erosion – gullies
vi)
Loss of riprap
vii)
etc
University of Nairobi –FCE 511 Geotechnical Engineering IV
- 82Spillway
i)
ii)
iii)
iv)
v)
vi)
Introduction to dam design
You might wish to recalculate the adequacy of the spillway. This topic is
covered separately under hydrology
Check field indicators of adequacy of the spillway – water marks
Blockages
Is the gear control working
Structural failures in the concrete
Note any cracks
Gauge house
i)
Are the instruments in good working order
ii)
Have they been vandalized
Reservoir area
i)
Assess the siltation
ii)
Assess the conservation measures being undertaken in the neighborhood of the
dam
iii)
What is the state of the fence of the reservoir fence for the fenced reservoirs?
AOB
3.4
Examples of earth dams in Kenya
In general the dam axis should be chosen in such a way as the material required for the
embankment is minimal while getting the maximum storage. Usually this is so where the
contours are narrowing downstream of a wide valley. The dam axis should be designed as
straight as possible unless the topographical features dictate otherwise
The height of the embankment should be determined in order to achieve the desired
storage with an increased gross freeboard. The gross freeboard is the height between the
spillway crest and the embankment crest and takes account of the design flood and the wave
height
The crest width should be such that earthmoving equipment can be able to work on
the crest. In many cases a road should cap the embankment. In any case a minimum width
of four meters should be observed.
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Introduction to dam design
Slope protection
4 Core
Depth
(m)
Side slopes
Core slope protection
5. Foundations
Soil type
6. Reservoir
Fetch
(m)
Depth
(m)
Area
(m2)
Capacity
(Mm3)
7. Bellmouth Spillway
Crest diameter (m)
Shaft diameter (m)
capacity
(m3/s)
8. Draw off system
Height of stand pipe(m)
Pipe Diameter
(mm)
Height of tower (m)
Table 3.2: Design statistics for Ndakaini dam
1Thika
Description
district
Dam name
Ndakaini (Thika)
District
Thika
Dam type
Zoned embankment
Designer/Engineer
Howard
&
1. Catchment area
Humpreys
Catchment
area 71
(km2Altitude
at dam site 2000
)
(masl)
Mean annual rainfall 1500
(mm)
General soil types
Grade IV to VI
3.
Embankment
Crest length
(m)
420
Crest height
(m)
65
Crest width
11
Bottom
width
(m)
(m)
Upstream side slope
3:1
Dow stream side slope
2.5:1
Freeboard
2
Embankment
(m)
2.5
volume(Mm3)
Crest
2000 masl
10000
culvert
intake
3
1
1
2.75
1
2
1:1.5u/s ,1:5d/s
Filter drains
Weathered
rock
4250
41
2900000
70
15
2
417
20
5500
70
2045 masl
11000
2.25
2025 masl
1
Downstreamshoulder
4000
5
Core
Drainage layers
2.25
1
2005 masl
1
70000
Full storage level 2041 masl
2.5
2030 masl
1
3000
Draw-off tower
3
1
2015 masl
1
1.5
3000
upstreamshoulder
3.5
Riprap
4000
Filter drain
2.25
Drainage blanket
1985
masl
1
5000
2.2
Original Ground level
Draw-off pipe
70 mdeep grout curtain
(a): Embankment details
University of Nairobi –FCE 511 Geotechnical Engineering IV
1
outlet
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Introduction to dam design
16mdiameter belmouth
embankment
5.5mpipe
(b) bellmouth spillway
Figure 3.2(a): Embankment details of Ndakaini Dam
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Introduction to dam design
Table 3.3: Design statistics for Kwa Tabitha dam, Kitui district
1. Description
Slope protection
Dam name
District
Dam purpose
Designer/Engineer
4
Core
Width
(m)
Kitui
5
Depth
(m)
Domestic water
supply
NWCPC
1
Side slopes
Kwa Tabitha
1. Catchment area
Altitude at dam site 1094
(masl)
Mean annual rainfall 720
(mm)
General soil types
Not available
3.
Embankment
Crest height
Crest width
(m)
Bottom
width
(m)
Hand
riprap
10.5
Core trench
(m3)
5. Foundations
1:2 u/s, 1:2d/s
volume 200
General soil type
Rock
6. Reservoir
Depth
(m)
7.1
7.
Spillway
5
Width at sill
(m)
38
15
Upstream side slope
Depth
(m)
3:1
2
Dow stream side slope
2:1
Freeboard
(m)
Embankment
1.5
12000
(m)
volume(m3)
25000
8.
Excavation
3
(m
) off system
Draw
Height of stand pipe(m)
10
Pipe Diameter
200
Crest
2300
500
1500
Gross freeboard
1
Drawoff
pipe
Protective
Hand placed gravel 300mm
riprap300mm
Grassing
Core
2
1
sand filter
Toe drain
5000
Figure 3.3: Embankment details of Kwa Tabitha Dam, Kitui district Dam
University of Nairobi –FCE 511 Geotechnical Engineering IV
(mm)
1101 masl
crest protection murram
1000
10mstand off pipe
5000
Normal water level ( 1100 masl)
3
placed
5000
1000
Cattle
trough
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Introduction to dam design
Table 3.4: Design statistics for Birica dam, Nyeri district
1 Description
4 Core
Dam name
Trench Width
Birica
District
Depth
Nyeri
Dam type
Side slopes
Embankment
Designer/Engineer
Core trench
NWCPC
(m3)
1. Catchment area
5. Foundations
Altitude at dam site
(masl)
Mean annual rainfall
(mm)
General soil types
3.
Embankment
Crest length
(m)
Crest height
(m)
Crest width
(m)
Bottom
width
(m)
Upstream side slope
Dow stream side slope
Freeboard
(m)
Embankment
3
)
volume(m
Slope protection
2161
1500
Not available
138
7
5
44
2.5:1
2.5:1
1.5
22500
Hand
riprap
1500
5000
placed
Rock
4
22000
50000
15
2
1800
2
200
2169 masl
5000
300mmProtective gravel
Normal water level
2.5
1
2.5
300mmHand placed riprap
1
Original ground level
2mlong stand-off pipe
7000
volume
Soil type
C’
6. Reservoir
Depth
(m)
Area
(m2)
Capacity
(m3)
7. Spillway
Width at sill
(m)
Depth
(m)
Excavation
(m3) off system
8. Draw
Height of stand pipe(m)
Pipe Diameter
(mm)
Crest
Gross freeboard
(m)
(m)
sand filter
8000
15000
Figure 3.4: Embankment details of Birica Dam, Nyeri district Dam
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Introduction to dam design
Table 3.5: Design statistics for Kwa Kasenga dam, Machakos district
1. Description
4 Core
Dam name
District
Kwa
Kasenga
Dam
Machakos
Trench Width
(m)
5
Depth
(m)
1
Dam type
Embankment
Side slopes
Designer/Engineer
NWCPC
1. Catchment area
2. Catchment
2
(km
)
3. Altitude
at
area 360000
dam site
1:2 u/s, 1:2d/s
Core trench
3
)
5. (m
Reservoir
volume 600
Fetch
(m)
Depth
(m)
(masl)
4. Mean annual rainfall 900
(mm)
5. General soil types
Not available
a.
Embankment
Crest length
(m)
Crest height
(m)
Crest width
(m)
Bottom
width
Area
(m )
Capacity
6.
Spillway
120
7
7
Length
41
Upstream side slope
3:1
7.
Excavation
3
(m
) off system
Draw
Dow stream side slope
2.5:1
Height of stand pipe(m)
2:1
Pipe Diameter
(m)
Freeboard
(m)
Embankment
3
volume(m
)
Slope protection
4.5
2
15000
3
(m )
7018
Width at sill
(m)
12.5
Depth
(m)
2
(m)
Not available
1800
Not available
(mm)
18000
Hand
riprap
placed
Crest
7000
Gross freeboard
1500
500
Normal water level
5000
3
1
300mm Protective gravel
Homogeneous embankment
300mm Hand placed riprap
2.5
7000
1
1000
sand filter
4000
5000
1000
Figure 3.5: Embankment details of Kwa Kasenga Dam, Machakos district
University of Nairobi –FCE 511 Geotechnical Engineering IV
Chapter Four : Site Investigation
4.1
Introduction
Site investigations are also referred to as soil exploration. It consists of investigating the
condition on which construction is planned. From site investigation it should be possible to
obtain information for the following geotechnical engineering activities
i.
ii.
iii.
iv.
v.
Design of new foundations
Modification of existing foundations
Location of materials of construction of roads, runways, etc
Identification of materials needed for the construction of pavement structures for
roads, runways etc
Identification of ground to be excavated in the construction of various facilities
including water pipe lines, building foundations, earthworks in cut areas etc
The site investigation should form a part of a coordinated chain of design from inception of
the project through preliminary to the final detailed design of a civil engineering project. It
should indeed continue post construction monitoring of the completed schemes. Because of
the diversity of civil engineering schemes a set of standard procedures is not possible for all
site investigations. The varying civil engineering schemes require a variety of options in
breadth and detail needed for the various schemes. The objectives for which a site
investigation is carried out also differ with various schemes. The main objectives of carrying
out a site investigation are now presented
i) Suitability of site for particular works
In the case of option of site for particular works a detailed site investigation should be able to
enable determination of the most suitable site. Thus it is possible to shift a bridge from one
location which would call for expensive deep foundations to one where ordinary shallow
foundations would be sufficient.
ii) Adequate and economic design
A site investigation leads to safe structures during and after construction. Additionally
sufficient information is obtained for quantifying the excavations needed in the preparation of
the bills of quantities. This should minimizes the possibility of cost overruns due to
unexpected ground conditions being met at construction time.
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Site Investigation
iii) Planning construction
By identifying different materials along the construction paths and their locations a
systematic procedure of carrying out the works is evolved. In the case of road works
materials from the cut areas are analyzed for use in the fill areas. It is then possible to
proceed with construction of the fills and cuts methodically with minimum haulages and
waste of materials.
iv) Prediction in changes in structure
Carefully and well executed site investigations should enable the prediction of the likely
settlement of structures under construction. Equally important is the ability to predict the
effect of excavations on the neighboring structures.
v) Safe structural design of large structures
Heavy modern structures require more detailed site investigations. Today we are seeing
higher buildings, larger bridges and installations sensitive to settlement. Structures and civil
engineering schemes are being put up very quickly. Immediate and consolidated settlement
is taking place when the works are commissioned. Further settlement takes place during the
useful life of the civil engineering installation. Accurate estimation of the settlement regime
is particularly important considering that clients are becoming more and more sensitive to the
performance of structures and the argument that cracks are minor and do not pose any
danger to the structure is no longer good.
4.1.2 Planning a site investigation
Table 4.1 shows a schematic way in which various activities with respect to site investigation
can be performed at various stages of a project. It is clear from the table that site
investigation should not be treated as an afterthought but rather should grow with the project
from conceptual initial design to eventual post construction period.
University of Nairobi –FCE 511 Geotechnical Engineering IV
Table 4.1 Stages of a site investigation
Phase
Stage
Main activity
Site
investigation
activity
SI Reports
Pre-construction
Conceptual
Initial design
Conceptual
design
Preliminary design
Design Alternatives
Detailed design
Detailed Site
Detailed investigations
Desk study of SI – -Boreholes
Review of existing -Trial pits etc
Define Scope of data Preliminary trial Laboratory and field
SI
pits
tests
Terms
of
reference
and i)
Preliminary SI
bid documents
investigation report
Detailed design report
ii) Cost estimate of SI -SI report
Construction
Supervision
construction
Post Construction
of Operation
Maintenance
Construction control
&
Field observations
– field densities
- field moisture contents -
-Performance
Monitoring
and
checking performance –
- pore water pressures
Settlement
Inclinations
As built SI report -
-Maintenance reports
-Performance reports
-Research reports
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4.2
Preliminary and detailed stage site investigations
4.2.1 Preliminary stage site investigations
This should lead to information needed for the design of the various alternatives at the
preliminary stage of the study. The activities in this stage can be summarized as follows:
i)
ii)
iii)
iv)
A study of any existing site investigation reports for the area or in the neighborhood
should form the basis of this stage of investigations.
A study of geographical a geological maps of the site in the case of large sites.
Topographical characteristics should lead to useful information such faulty areas.
Heavily forested areas are an indication of deep rooted top soils.
A site inspection of the existing buildings and any existing structures. Any signs of
distress which can be related to the settlement of the foundations. Any information
from archives, previous records held by the local authorities.
Inspection of the soil profiles, in cut areas, old used quarries. Structured questions to
local people with regard to the geotechnical information being sought yields
considerable information. Such questions are:
a)
b)
c)
What is the depth of the pit latrines in the area?
At what depth murram encountered?
At what depth was water struck?
v)
Aerial survey of the site could give useful information with regard to land formations
and soil profiles.
vi) Seismic refractions could be carried out at this stage of investigations. Usually a
specialist is needed to interpret the results.
vii) Preliminary trial pits
Geophysical methods
Geophysical methods involve sending of seismic or electrical waves through the ground. The
determination of the soil strata is based on the fact that the velocity or the resistance seismic
wave transmission or resistance to electrical flow differs with different rock types and soils.
The method allows the boundaries of the soils to be determined seismic refraction is
described below
Seismic refraction is conducted by having a source of seismic waves (Figure 4.1). The
seismic waves are induced by detonating a small explosive or by striking a metal plate hard.
Waves are subsequently emitted in all directions, through the air, and through the soil in all
directions. Seismic wave transducers called geophones are placed radially from the
epicenter. A circuit connects the geophones and the detonator for accurate determination of
time. A direct wave will reach the geophone first since it is the shortest distance covered.
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Site Investigation
When there is a dense stratum at depth a refracted wave will travel along the top of the bed
rock. As it travels it leaks energy to the surface which can be picked by the geophone.
Seismic source
Geophones
Figure 4.1 Seismic refraction – arrangement of equipment
Time
For short distances the direct waves reach the geophones first. For longer distances
the refracted wave reaches first though the distances is longer than t he surface direct
distance. This is so because the speed of the wave in the dense material is higher than that in
the overburden material of less density. The geophone has a mechanism which records the
first wave and ignores the others. This enables a plot of arrival time versus the distance.
The first section of the graph represents the direct wave measurements while the second
section represents the refracted wave measurements (
Figure 4.2). The inverse of these curves are the velocities of the seismic waves. The
general types of the rocks are determined by geophysics from the knowledge of velocity
versus rock type. It is also used in the determination of depth to water table and thicknesses
of multiple strata. The depth D to the bedrock can be estimated from the formula.
d
Distance
Figure 4.2 Time versus distance for seismic waves
√
4.2.2 Detailed stage site investigations
At this stage the aim is to obtain data for use in the final design of the works. The
investigation is carried out by use of trial pits, sounding and boring. The extent of the use of
these methods depends on the type of the project at hand and the geotechnical parameters
being sought.
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Site Investigation
The trial pits
The pit and shaft technique supplies the most detailed and reliable data on he existing soil
conditions. Once the trial pit has been dug stratification of the soil should be done usually in
the field. In addition as much information should be recorded. This information includes
i.
ii.
iii.
iv.
Depth to ground water table.
Field assessment of the bearing capacity.
Depth of the various strata encountered in the trial pit.
The encountered soils should classified by visual inspection
a.
b.
c.
v.
Coarse grained soils should be described with adjectives such as angular,
rounded with traces of fines etc
Fine grained soils should be studied to indicate whether they are loamy, of low
plasticity, whether they are sandy clays etc
All soils should be described indicating their color and odour if any. Decaying
organic matter if encountered should be mentioned.
Obtain undisturbed samples when you can for the different layers of strata
encountered. These samples can then be taken to the laboratory for tests
For large sites the pits should then be surveyed and located in a grid system for incorporation
into the site investigation report.
Sounding tests
These are basically are penetration tests carried out to supplementing trial pits and borings.
The penetration resistance is measured and related to the bearing capacity. They are widely
used in site investigations. They consist of the cone penetrometer already presented in
chapter 1. The other commonly used penetration equipment is the dynamic cone
penetrometer used in the estimation of the California bearing ratio (CBR) of road pavement
layers. This enables the design of the pavement layers to be carried out
Boring methods
When a deep stratum has to be investigated it will usually be necessary to perform boring
operations to ascertain the strata below the ground to be used in the support of the proposed
structures. Several boring methods are available and are summarized as follows
Percussion drilling consists of a derrick, a power unit and a winch carrying a light steel cable
which passes thorough a pulley. The unit can be towed by a vehicle after the assembly is
folded. The assembly drops a chisel on the ground and strata being drilled
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Site Investigation
Rod
Chisel
Figure 4.3 Schematic presentation of a drilling chisel
The excavation is effected by the drilling chisel. The drilling rods provide the necessary
weight for the penetration the strata. Further weight may be added when need arises. The
winch raises and lowers the chisel and its attachments
Below the water table the loosened soil forms slurry. Above the water table water is
introduced to form the slurry. Periodically the slurry is bailed out by a shell or a bailer to
make progress into the soil. In boreholes which are liable to collapse the borehole must be
cased. In some cases the casings slide on their own weight. On completion of the job, the
casing is jacked out.
Percussion drilling is usually done in diameters of 150mm to 300mm. the borehole
depth investigated by this drilling method can be up to 50 to 60 metres. This method of
drilling can be done on virtually all types of soils including those with boulders and cobbles.
The rig is versatile enough to place mechanical augers and penetrating testing equipments at
appropriate depths.
Power operated augers are usually on vehicles. Downward pressure is applied by pressure
or dead weight. The augurs are 75-300mm diameters. Augers are usually used in self
supporting soils. Casing is usually not needed since the augers have to be removed before
driving. In full flight augers the rod and the helix cover the entire length being investigated.
The augur is then brought up. The soil is ejected by reverse rotation. The likely hood of soil
from different strata being mixed up is very high. In the short flight augur the auger is
advanced into the soil and then raised. The soil is also ejected by reverse rotation.
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Full flight augur
Site Investigation
Short flight augur
Figure 4.4 full flight and short flight augurs
The continuous flight augurs are sometimes fitted with a hollow stem which is plugged
during the drilling operations. When samples are needed the plug and the rods are removed
and a sampler is introduced for the recovery of a sample. The sample may be undisturbed
depending on the sampler utilized. The flight augurs are not suitable for use in loose soils
which are likely to collapse as the augur is inserted and removed from the hole.
Hand and portable augers are usually operated by persons by turning the handle of the
augur. The hand augers are typically of 75 – 300mm diameters. The soil is locked in the
auger and frequent removal is needed to ensure that the augur does not get stack in the soil.
Undisturbed samples may be obtained by introduction of small diameter tubes which are
hammered into the strata under investigation. This method is suitable for self supporting
soils. It is not possible to penetrate coarse granular soils.
Figure 4.5 schematic representation of a hand augur
Wash boring is a method of boring where water is pumped through boring rods and released
through narrow holes in the chisel attached at eth lower ends of the boring arrangement
(Figure ****).
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Site Investigation
Water from pump
Tiller
To sump
Drilling bit
Figure 4.6 schematic representation of a hand augur
In this method the soil is loosened and broken by water jet. This is aided by the up an down
movements of the chisel. An attachment to the rods called a tiller enable the rotation on the
drilling bit. The drilling winch is able to raise and lower the chisel and hence get the
chopping action of the chisel.
This method is suitable for most soils but progress is slow if the particles of coarse
gravel larger particles are present. The accurate identification of the soil types is difficult.
The method cannot be used to recover soil samples for testing. However tube samplers can
be advanced into the borehole for obtaining relatively undisturbed samples.
Rotary drilling is done by use of drilling bits that cuts and grinds the subsoil or rock at the
bottom of the borehole. Water is usually pumped down hollow rods passing under pressure
through to the drilling tools. This cools and lubricates the bits. The fluid also provides
support for the borehole where there is no casing.
Two methods of rotary drilling are available. The first is open drilling where the soils
and rocks are broken within the diameter of the hole. Subsequently the tubes are removed and
tube samplers and testing continues below the borehole. This advances the drilling. The
second method is known as core drilling and involves creation of an annular hole in the
material and intact rock enters the drilling core. This advances the drilling and enables
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Site Investigation
samples to be retrieved from the borehole. The sample is then subjected to immediate field
description and taken to the laboratory for various tests. Typical core diameters range from
41mm to 165mm. The method is fast, but in large gravelly soils the speed is slowed by
rotation of the bit without advancement into the ground.
4.2.3 Sampling
Disturbed samples
Disturbed samples are recovered from trial pits and along drilling tools where there is no
attempt to retain the soil constituents. Disturbed samples should however be collected
carefully and placed in airtight tins or jars or in plastic sampling bags. The samples should be
labeled to give the borehole or trial pit identification number, depth of recovery and field
description should be done. The disturbed samples are used for identification tests namely
Field moisture content, PI, grading, compaction and CBR.
Undisturbed sample – cohesive soils
Undisturbed samples are recovered from trial pits and along drilling tools where there is an
attempt to retain the soil constituents. Such a sample is taken in an airtight container with
wax at both ends to prevent moisture from escaping during transportation to the laboratory.
In trial pits the samples can be obtained by pressing a sampler into the ground at the
appropriate depth. The sampler is typically 100mm diameter by 150mm long. In the hand
augur a 38mm sampling tube with a length of 200mm is fitted to the rod after the removal of
the augur. The tube is pressed into the soil and given half a turn to break the soil. The
sampler is then removed and the ends are waxed. In boring rigs a 105mm diameter sampler is
introduced to the borehole to recover a 100mm diameter sample. The sample is usually
381mm long and is fitted with a cutting shoe of about 110mm diameter. The sample is driven
by a falling weight. Any entrapped air or water is expelled from the top through a non return
valve. For soft clays thin walled samplers are preferred to minimize disturbance.
Inevitably there will be some disturbance in the process of retrieving soil samples
from the ground. The least disturbance is for shoes samples cut from the floor of trial pits.
Sample tubes, inserted by pressing, jacking or steady hammering produce some form of
disturbance depending on the thickness of the sampler walls. The degree of disturbance is
related to the area ratio of the sampler tube as given by Equation ****** In general good
samplers have and area ratio not exceeding 25%. Area ratios less than 10% are very good
and are used for very sensitive soils.
x100%
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Site Investigation
De
De
Di
Di
Sampler tubes
Sampler tubes fitted
with a cutting shoe
Figure 4.7 Typical sample tubes
Undisturbed sample – cohesionless soils
Various methods have been employed to obtain undisturbed sand samples. These include
freezing, chemical application, and use of compressed air (Smith and Smith, 1998).
Whatever method is employed eventual disturbance occurs as the soil is transported to the
laboratory for testing. In light of these difficulties it is prudent to assess the engineering
properties of cohesionless soils through field testing such as penetration.
Quality class for soil sampling
Table ** below based on Rowe (1972) shows the quality classes for soil samples obtained
from various site investigation operations.
Table 4.2 Quality class for soil sampling
Quality
class
5
4
3
2
1
Method of sampling
Material brought up by drilling tools an no attempt is
made to retain all the soil constituents
As for 5 but all soil constituents are retained as far as
possible. Bulk an jar samples. Plastic bag samples
Pressed or driven thin or thick walled samplers with
water balance in very permeable soils
Use of sample
Rough sequence of
strata
Sequence of strata and
remolded properties
As
above
and
examination of soil
fabric
As for class 3 above but with water balance all the As 3 and γ, n, mv, cu,
time
c’ θ’
Thin walled piston samplers with water balance
As 2 and cv and k
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Site Investigation
Borehole logs
Borehole logs summarizes all the laboratory an field tests carried out on samples representing
the various strata encountered in the boring operations. All ground conditions encountered at
the site are also included. The log enables a rapid accurate assessment of the soil profile on a
vertical scale. The details of the various strata encountered including all their geological
formation details which can be inferred are given. The details captured should include the
depth to which ground water was encountered. The description is based on particle
distribution and plasticity based visual inspection and feel. Soil color should also be
recorded.
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Courtesy of Norken Engineering Consultants
Figure 4.8 Borehole logs
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Site Investigation
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Site Investigation
4.2.4 Scope of Site Investigation
Spacing of the trial pits and or boreholes
The scope of site investigation is dependent on the effect of the construction on the ground.
The scope should be commensurate with the needed geotechnical parameters. Table 4.3
shows the suggested minimum number of borings for the various structures.
Table 4.3 Recommended spacing of investigation trial pits and boreholes
Project
Multistory
1 to 2 storeys
Bridge piers
abutments
Type of soil/Distance between borings
Uniform
Average
Erratic
45
30
15
60
30
15
and 30
30
15
Minimum no
4
3
1 – 2 per unit
For highways and runways during preliminary design the subgrade soils along the proposed
alignment should be sampled at 1000metres and the samples should be tested to establish the
in-situ CBR, grading and plasticity of the materials. At this stage the material site should be
investigated at 60 meter intervals. In the detailed stage the subgrade is sampled at 500meters
while the material sites are sampled at 30metres.
Depth of investigation
The depth should be such as to capture the geotechnical information needed for the design of
the facility. Equally important is to capture the information needed in the quantification of
the bill of quantities to ensure an accurate specification of the works is carried out. The
recommended depths below the formation of investigation for the various civil engineering
schemes is shown on Table 4.4 based on Figure 4.9 below.
Table 4.4 Depth of investigation
Project
Column foundations
Raft foundations
Bridge piers and
abutments
Earthworks in fill for
highways
Earthworks in cut
highways
Pipe works
Depth
1.5B-3B
1.5B
1.5B-3B
In rock
1.5-3m
1.5-3m
1.5-3m
Parameters to be established
C, θ, N, RQD,TCR
C, θ, N. RQD,TCR
C, θ, N, RQD,TCR
0.5L
0.50m
0.5H
0.50m
D
0
PI, CBR for fill material
Strength of support
Establish the type of excavated
material and strength of support
Investigate type of excavated
material and strength of support
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Site Investigation
Raft foundations
Column
foundations
B
Piled foundations
B
L
B
H
Retaining walls
B
a)
Structural foundations
L
L
In cut H
In fill
b)
Highway earthworks
D
c)
Pipe works
Figure 4.9 Scope of foundation investigations
4.2.5 Site Investigation Reports
List of suitable headings
Title page
Gives the title of the project at a glance
Abstract
The abstract should be approximately 200 words. It is a very important element of the
project and should be prepared with care. It must convey the essence of the site investigation
and all the important findings without ambiguity.
List of contents
Guides the reader to the various chapters
Field work
A brief and complete description of what was done in the field. Boreholes, and trial pits
performed, field testing etc. Actual procedures of standard tests need not be repeated. A
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Site Investigation
mention of the tests performed is sufficient. New procedures and peculiar fieldwork should
be explained.
Laboratory work
A brief and complete description of what was done in the laboratory work carried out . as in
the case of field testing actual procedures of standard tests need not be repeated. A mention
of the tests performed is sufficient. New procedures and peculiar laboratory equipment and
procedures should however be explained
Site description and geology
An engineering summary of the nature of the site an its geology, including aspects such
excavated areas and what was found, stability of natural slopes, drainage etc
Engineering properties of soils an rocks
A summary of the results of field and laboratory tests and other observations made at the site
Discussion
A reasoned discussion of what design and construction problems are likely to be encountered
in relation to the site and its geological situations.
Recommendations and conclusions
A brief but clear statement of the recommended geotechnical parameters investigated. The
treatment of the various aspects of design should come out clearly and without doubt. Values
of use in design and construction should be summarized viz, allowable bearing capacity,
estimated settlement, suitable types of foundations, construction requirements namely
grouting, compaction etc
References
A list of the books, papers, referred to in the work
Appendices
Appendix A – should contain site plan, borehole logs, photographs, etc
Appendix B – should contain tables of results of field and laboratory test those not included
in Appendix A
Appendix C – Any special or unusual test procedures adopted in the investigation
References:
Craig FR, 1987, Soil mechanics, Van Nostrand Reinhold (International) London
Bowles JE , 1982, Foundation Engineering, McGraw-Hill international book company,
Tokyo.
Tomlinson MJ and Boorman R (1986), Foundation and construction, Longman scientific and
technical, England
Franklin JA and Dussealt MB (1989) Rock Engineering, McGraw-Hill international editions,
London
Chen FH (1975) Foundations on expansive soils, Elsevier scientific Publishing Company
University of Nairobi –FCE 511 Geotechnical Engineering IV
Department of Civil and Construction Engineering
Geotechnical Engineering (FCE 511)
Teaching notes
By Sixtus Kinyua Mwea
2015
University of Nairobi –FCE 511 Geotechnical Engineering IV
Syllabus
FCE 511 - Geotechnical Engineering III
Foundations:
Shallow Foundations
Introduction. Foundation loading intensities. Bearing capacity, (ultimate, safe, gross and
allowable). Influence of ground water table, sloping ground, inclined and eccentric loads on
allowable bearing capacity. Design of shallow foundations for shear strength and settlements.
Examples of foundation design (e.g. strips, pad), combined footings, raft footings.
Piled Foundation
Types of piles driven and bored pile, friction and end bearing pile. Design of piles by soil
mechanics methods, end bearing, skin friction and ultimate bearing resistance. Piles in sands.
Piles in cohesive soils - total and effective stress analysis. Design from pile tests data.
End bearing piles on rock. Settlement of piles. Dynamic formula. Negative skin friction. Pile
groups - bearing capacity in cohesive and cohesionless soils.
Introduction to Earth Dams
Design of earth embankment - homogenous and zoned dams. Definitions e.g. fetch, water
spread, shell free board etc. Factors influencing site selection. Spillways. Settlements of
embankments. Protection of upstream and downstream slopes.
Site Investigations
Introduction, purpose of Site Investigation, organization of Site investigation for different types
of structures e.g. buildings, irrigation or water supply projects, highways and airport pavements,
etc. Methods of Investigation. Sampling. Borehole logs. Geophysical methods. Geotechnical
reports.
University of Nairobi –FCE 511 Geotechnical Engineering IV
i
Table of contents
Syllabus ................................................................................................................................... i
Chapter one ................................................................................................................................ 1
Shallow foundations ................................................................................................................... 1
1.1
Types of foundations .................................................................................................. 1
1.2
Introduction to shallow foundations ........................................................................... 1
1.2
Bearing capacity of soils ............................................................................................ 2
1.2.1
Bearing capacity terms ......................................................................................... 2
1.2.2
Ultimate bearing capacity..................................................................................... 3
1.2.3
The net foundation pressure ............................................................................... 12
1.2.4
Allowable bearing pressure ................................................................................ 13
1.2.5
Field methods for the determination of bearing capacity of soils ...................... 14
1.2.6
Presumed bearing capacity of soils and rocks .................................................... 23
1.3
Proportioning of shallow foundations ...................................................................... 24
1.3.1
Contact pressure distribution .............................................................................. 24
1.3.1
Proportioning the foundations ............................................................................ 25
1.3.2
General consideration in the selection of the foundation depth ......................... 34
1.3.3
Foundations for common buildings.................................................................... 35
1.4
Foundations for difficult soils .................................................................................. 36
1.4.1
Foundations on expansive clays ......................................................................... 36
1.4.2
Foundations on loose sands ................................................................................ 41
1.5
Tutorial examples on chapter one ............................................................................ 43
Chapter two .............................................................................................................................. 45
Deep Foundations ..................................................................................................................... 45
2.1
Pile foundations ........................................................................................................ 45
2.1.1
Introduction ........................................................................................................ 45
2.1.2
Classification of Piles by materials and construction......................................... 46
2.1.3
Driven piles ........................................................................................................ 48
2.1.4
Bored piles.......................................................................................................... 51
2.1.5
Determination of pile load carrying capacity ..................................................... 53
2.1.6
Determination of load carrying capacity dynamic methods............................... 59
2.1.6
Determination of load carrying capacity pile testing ......................................... 61
2.1.7
Negative skin friction ......................................................................................... 62
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2.1.8
2.2
Pile groups .......................................................................................................... 64
Drilled piers and Caisson Foundations..................................................................... 66
2.2.1
Drilled piers ........................................................................................................ 66
2.2.2
Caisson Foundations .......................................................................................... 66
2.4
Examples of Piling Schemes .................................................................................... 71
2.5
Tutorial examples on chapter two ............................................................................ 71
Chapter Three ........................................................................................................................... 73
Introduction to Earth Dams ...................................................................................................... 73
3.1
Introduction .............................................................................................................. 73
3.2
Selection of type of earth dam .................................................................................. 74
3.2.1
Diaphragm types ................................................................................................ 74
3.2.2
Homogenous types ............................................................................................. 75
Zoned types ..................................................................................................................... 75
3.2
Design Principles ..................................................................................................... 76
3.3.1
Foundation design .............................................................................................. 76
3.3.2
Embankment Design .......................................................................................... 79
3.3
Inspection of existing dams ...................................................................................... 81
3.4
Examples of earth dams in Kenya ............................................................................ 82
Chapter Four ............................................................................................................................. 88
Site Investigation ...................................................................................................................... 88
4.1
Introduction .............................................................................................................. 88
4.1.2
4.2
Planning a site investigation ............................................................................... 89
Preliminary and detailed stage site investigations .................................................... 91
4.2.1
Preliminary stage site investigations .................................................................. 91
4.2.2
Detailed stage site investigations ....................................................................... 92
4.2.3
Sampling............................................................................................................. 97
4.2.4
Scope of Site Investigation ............................................................................... 101
4.2.5
Site Investigation Reports ................................................................................ 102
References: ............................................................................................................................. 103
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Geotechnical Engineering IV
Week
Introduction
1 2
3
4
5
6
7
8
9
10 11 12 13 14 15
Shallow foundations
Foundation intensities
Bearing Capacity
Factors that influece bearing capacity
Design of shallow foundations
Piled foundations
Types of piles
Types of piles
Driven piles
Bored piles
Pile load capacity
Settlement of piles
Negative skin friction
Pile groups
Introduction to earth dams
Definitions (fetch water
freeboard)
Design of earth embankment
Site selection
Spillways
Settlement of embankments
Protection of slopes
spread,
Continuous Assessment Test
Site investigation
Introduction
Purpose of site investigation
Organization of site investigation
SI for different schemes
Methods for site investigation
Geotechnical reports
Revision and tutorials
Main examinations
Target dates
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Chapter one: - Shallow foundations
1.1
Types of foundations
Foundations that are encountered in practice may be classified into two broad categories
namely shallow and deep foundations. Under shallow foundations the following categories
are usually encountered:a)
b)
Strip foundations for wall and closely spaced columns
Spread or isolated footings for individual columns. In this category it is usual
to consider combined foundations for two or three closely spaced columns as
spread or isolated footings
c)
Raft foundations covering large sections of the foundation area
The design and construction of shallow foundations is dealt with in this chapter.
Under deep foundations the following two types of foundations are encountered :a)
b)
Piles
Caissons
The design and construction of deep foundations is dealt with in the next chapter.
In the selection of the foundations to adopt for a structure it is usually necessary to
consider the function of the structure, its loads, the subsurface conditions and the cost of the
foundation being adopted in comparison to other possible types of foundations.
1.2
Introduction to shallow foundations
The foundation is the part of the structure that transmits the loads directly to the underlying
soil. If the soil is sufficiently strong it is possible to use shallow foundation. On the other
hand if the soil is not strong enough the foundation is taken deeper into the ground and is
referred to as a deep foundation. A definition which sometimes conflicts with the definition
of the shallow foundation defines a shallow foundation as one whose depth is less or equal to
its least width. The foundation must satisfy two fundamental requirements:-
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Shallow foundations
1. The factor of safety against shear failure must be adequate. A value of 3 to 5 is
usually specified.
2. The settlement of the foundation should be tolerable and in particular differential
settlement should not cause any unacceptable damage o interfere with the function of
the structure.
3. The allowable bearing capacity is defined as the pressure which may be applied to the
soil to enable the two fundamental conditions to be satisfied
The damage being mitigated in the design of the structures can be classified as architectural,
functional or structural. In the case of framed structures settlement damage is usually
confined to the cladding and finishes (architectural damage). It is usual to expect a certain
amount of damage. What is critical is to ensure that the damage to the services is limited.
Angular distortion limits were proposed by Craig (1987) and are shown on Table 1.1. In
general the limiting angular distortion to prevent damage is 1/300. For individual footings
this translates to a maximum settlement of about 50mm in sand and 75mm in clay. An
accurate damage criterion is to limit the tensile strain at which the cracking occurs. The
concept of tensile strain should be used in analysis using an idealization of the structure and
the foundation in elastic strain analysis when the fundamental properties of the foundations
are known.
Table 1.1 Angular distortion limits
1/150
1/250
1/300
1/500
1/600
1/750
Structural damage of general buildings may be expected
Tilting of high rigid buildings may be visible
Cracks in panel walls expected
Difficulties with overhead cranes
Limit for buildings in which cracking is not permissible
Overstressing of structural frames without diagonals
Difficulties with machinery sensitive to settlement
The design of the foundations is usually a two process exercise. The first is to determine the
allowable bearing of the soil while the second is to size the foundation on the design strata
based on the allowable bearing capacity. These two parts are now discussed.
1.2
Bearing capacity of soils
1.2.1 Bearing capacity terms
The following terms are used in bearing capacity problems
Ultimate bearing capacity is the value of the average contact pressure between the foundation
and the soil which will produce shear failure in the soil.
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Shallow foundations
The net foundation pressure is the increase in the pressure at the foundation level due to the
structure loads
The safe net foundation pressure is the net foundation pressure divided by a suitable factor of
safety
Allowable bearing pressure is the maximum allowable net loading intensity on the soil
allowing for both shear and settlement effects.
1.2.2 Ultimate bearing capacity
If a load is increased at the foundation level, shear failure would take place in the foundation
at a load which can be referred to as failure load. The resulting pressure at the base of the
foundation is known as the ultimate bearing capacity of the soil
Three distinct modes of failure have been identified and these are illustrated in
Figure 1.1 in the case of strip footing. As the pressure increases on the foundation layer the
state of plastic equilibrium is reached initially in the soil around the edges of the footing and
then spreads downwards and outwards. Ultimately the state of plastic equilibrium is reached
throughout above the failure surfaces. The soil around the footing heaves on both sides. At
the moment of failure one side continues to settle at a higher rate and the strip footing tilts.
This behavior is exhibited by soils of low compressibility (Figure 1.1a). . Local shear failure
is characterized by local development of plastic conditions usually below the foundation. The
plastic conditions do not reach the surface and only slight heaving is expected. This kind of
failure is expected with soils of high compressibility and is associated with large settlements
(Figure 1.1b). These soils include dense and stiff soils. Punching shear occurs when shearing
takes place directly below the footing under compression from load. No heaving is of the
ground is expected by the side of the footing. Large settlements are characteristics of this
mode of failure and are typical of soils of high compressibility and foundations at
considerable depth (Figure 1.1c). In general the mode of failure will depend of the
compressibility of the soil and the depth of the foundation.
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a)
Shallow foundations
General shear failure
Pressure
a
Settlement
b)
local shear failure
c
b
c) Punching shear failure
Figure 1.1 Modes of failure of foundations
Bearing capacity by use of earth pressure analogy
The earth pressure analogy can be explained by consideration of a strip footing on a
cohesionless soil as shown on Figure 1.3
q
p
γD
Figure 1. 2 Pressure below a strip footing
The vertical pressure is q which is a result of the structure loads. By use of Rankine active
pressure theory, a lateral pressure p holds the soil in equilibrium below the foundation. For
particles just beyond the edge of the foundation the lateral pressure is more than the vertical
pressure γD resulting from the overburden. The vertical pressure γD is the minor principle
stress and p is the principal stress. By use of the Rankine earth pressure theory Equations 1.1
through 1.3 can be deduced.
p q(1sin)/(1sin) (inside the foundation)
1. 1
p D(1sin) /(1sin) (outside the foundation)
q D((1sin)/(1sin))2 (ultimate bearing capacity) 1. 3
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Shallow foundations
For a cohesionless soil the bearing capacity is dependent on the overburden and equals to
zero for a foundation on the ground surface. Bells development for a c-υ is given in Equation
1.4
q D((1si)/(1sin))2 2c ((1sin)/(1sin)3 2c (1sin)/(1sin)
.1. 4
For a purely υ =0 soil the ultimate bearing capacity is given by Equation 1.5
q D 4c
1. 5
Bearing capacity by use slip circle analogy
The slip circle analogy can be explained by consideration of a strip footing on a cohesive soil
as shown on Figure 1.3
B
q
D
O
B
πB
Figure 1. 3 A slip circle analogy of a strip footing
The foundation is assumed to fail by rotation about a slip surface of radius equal to the width
of the base B and at the edge of the foundation O. At ultimate conditions the disturbing
moment (Md) is given by Equation 1.6
Md q*L*B* B
2
1. 6
The resisting moment (Mr) about O is a summation of the resistance due to the cohesion on
the cylindrical surface, on the vertical surface and the weight of the overburden as given in
Equation 1.7
Mr cLB2 CDLB DLB
2
2
1. 7
At ultimate conditions the disturbing moment is equal to the resisting moment and the
ultimate bearing Equation for a υ = 0 soil is given by Equation 1.8
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Shallow foundations
q 6.28c(1 0.32D 0.16D)
B
c
1. 8
Plastic theory failure
A suitable failure under a strip footing is shown on Figure 1.2. The footing of width b and
infinite length carries a uniform pressure of magnitude qf. The shear strength parameters for
the soil are c and υ. The unit weight of the soil is assumed to be zero. At ultimate bearing
capacity the soil is pushed downwards into the soil mass producing a state of plastic
equilibrium in the form of an active Rankine zone below the footing where the angles ABC
and BAC are each 45+υ/2. The zone ABC resists movement and is intact with the base. It
suffers no much deformation. The downward movement of the wedge ABC forces the
adjoining soil to move sideways. Passive Rankine zones ADE and GBF are developed and
angles AEF and BFG are 45-υ/2. these zones confine the movement of the wedge EDA and
BGF. The transition between the downward movement of the wedge ABC and the lateral
movement of the wedge EDA and BGF takes place through zones of radial shear ACD and
BCG. The surfaces DC and CG are logarithmic spirals. The soil above EDCGF is in a state
of plastic equilibrium while the rest of the soil is in state of elastic equilibrium.
qf
qo
A
B
45+φ/2
F
45-φ/2
E
D
C
G
Figure 1. 4 Failure under a strip footing
Using plastic theory the ultimate bearing capacity below a strip footing on a surface of a
weightless soil is given by Equation 1.9. This is for undrained condition where υu = 0
qf (2)cu 5.14cu
1.9
In general the foundation is located at a depth and imposes a surcharge qo = γD. The weight
of the surcharge and the pressure of the foundation produce stresses on the moving masses of
soil at plastic conditions.
The ultimate bearing capacity of the soil under shallow strip footing can be
expressed by the following general equation suggested by Terzaghi.
qf 0.5BN CNc DNq
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Shallow foundations
Nγ, Nc and Nq are bearing capacity factors which depend on the values of υ. Nγ
represents the contribution to the bearing capacity resulting from the self weight of the soil.
Nc is the contribution due to the constant component of the shear strength and Nq is the
contribution of the surcharge pressure. Values of Nγ, Nc and Nq can be obtained from
Equations 1.11 through 1.13 the values for Nc and Nq were suggested by Meyerhof (1955)
while the values of Nγ, were suggested by Hansen (1970) These values are plotted in terms of
υ in Figure 1. 5.
Nc (Nq 1)cot
Nq tan2 (45/ 2)e tan
N 1.5(Nq 1)tan
1. 11
1. 12
1. 13
Values of Nc, Nq, Nγ
Nq
100
Nγ
Nc
10
1
0
10
20
30
40
φ - Degrees
Figure 1. 5 Bearing capacity factors for shallow foundations
Bearing capacity for square, round and rectangular foundations
The problem involves extending what is basically a two dimension problem in a strip footing
to a three dimension problem in other foundation shapes. The bearing capacity factors for
square and round foundations are shown on Equations 1.14 and 1.15 respectively.
q 0.4BN 1.3cNc DNq
q 0.3BN 1.3cNc DNq
University of Nairobi –FCE 511 Geotechnical Engineering IV
1. 14
1. 15
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Shallow foundations
The factors for rectangular footing are an interpolation of the square and the strip footing and
are shown on Equation 1.16
q 0.5BN (10.2B/ L) cNc (10.3B/ L) DNq
1. 16
It showed be noted that the values of the bearing capacity factors are very sensitive to the
values of shear strength parameters c and υ. Due consideration should therefore be given to
the degree of accuracy of these values. In general the following observations have been made
a)
b)
c)
d)
e)
In cohesive soils the contribution of cohesion c to the bearing capacity dominates
The depth factor dominates for cohesionless soils
The base factor is usually neglected for values of B less than 4 meters
A footing at the surface has no bearing capacity if Nγ is neglected
The equations are applicable to uniform soils and in the case of stratified soils an
engineering judgment is always required.
Skempton’s values of Nc
Skempton (1951) showed that for a cohesive soil (υ =0) the value of Nc increases with the
value of foundation depth D. He suggested that the values of Nc applicable to circular, square
and strip foundations are given in Figure 1.6. The value of the rectangular footings of
dimensions BxL (where B<L) is the value of a square footing multiplied by (0.84+0.16B/L).
10
9
Nc
8
7
Nc (Strip)
6
Nc (Circular or
Square)
5
4
0
1
2
3
D/B
Figure 1.6 Skempton’s values of Nc for a φ =0 soil
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Shallow foundations
Eccentric and Inclined loading
Eccentric and inclined loadings have an effect of reducing the foundation bearing capacity.
In the case of a foundation with a vertical load such that the eccentricities are e b and el
(Figure 1.7 ) the effective foundation dimensions are shown as B’ and L’ The resulting load
is distributed over the effective foundation dimensions. The values of B’ and L’ are given in
Equations 1.17 and 1.18
B' B2eB
L' L2eL
1. 17
1. 18
In the case of inclined load (Figure 1.8) on a width B and inclination the effective
foundation width is B-2e. In addition the bearing capacity factors are multiplied by the
inclination factors shown on Equations 1.19 and 1.20
ic iq (1/90o )2
i (1/)2
1)
1. 19
1. 20
The base of a long retaining wall is 3m wide and is 1m below the ground in front
of the retaining wall. The water table is well below the base level. The vertical
and horizontal components of the base are 282kN/m and 102kN/m respectively.
The eccentricity of the base reaction is 0.36m. The appropriate shear strength
parameters are c’= 0 and υ’ = 35o. The unit weight of soil is 18kN/m3.
Determine the factor of safety against shear failure
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Shallow foundations
282kN/m
1m
282kN/
1.14m
102kN/m
.36m m1.5m
Solution
For υ’ = 35o, Nγ = 41 and Nq = 33
The angle of the inclination to the vertical α = tan -1 (102/282) = 20o hence the inclination
factors according to Meyerhof are
The ultimate bearing capacity is given by
The factor of safety
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Shallow foundations
An alternative approach in the case of inclined loads is to use the empirical formula shown on
Equation 1.21
L’
Y
B’
B
eB
X
eL
L
Figure 1. 7 Effective dimensions for pads subjected to eccentric loads
PV / Pav PH / Pah 1
1. 21
Where Pv is the vertical component of the inclined load and PH is the horizontal
component of the inclined load. Pva is the allowable vertical load and PHa is the horizontal
load (a fraction of the available passive resistance).
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Shallow foundations
α
Pv
P
PH
e
B
Figure 1.8. Foundation with inclined load
1.2.3 The net foundation pressure
The actual pressure on the soil due to the weight of the structure is called the total foundation
pressure q. The net foundation pressure qnet is the increase in the pressure at the foundation
level. This is the total foundation pressure less the effective weight of the soil permanently
removed during excavation and is given in Equation 1.22
qnett q D
1. 22
For a strip footing the net foundation pressure is shown on equation 1.23
qnett 0.5BN cNc D(Nq1)
1. 23
The safe net bearing pressure (qsafe) is the net bearing pressure factored by an appropriate
factor of safety as shown on Equation 1.24
qsafeqnett/ FOS
1. 24
It is usual to use conservative factors of F usually between 3 and 5. Due to uncertainties in
the determination of the strength parameters
and determination of the of the service load,
for comparison the following factors of safety (Table 1.2) are used in other geotechnical
works
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Shallow foundations
Table 1. 2 Typical factor of safety values for geotechnical works
Failure mode
Shear
Shear
Shear
Seepage
Shear
Type of works
Earthworks
Retaining walls
Sheet piles
Uplift
Bearing Capacity
FOS
1.2-1.6
1.5-2.0
1.2-1.5
1.5-2.5
3-5
Effect of ground water
Water table below the foundation level
If the water table is at a depth not less than B below the foundation level the expression for
the net ultimate bearing capacity is given in Equation 1.23 above. However the when the
water table rises to a depth less than B below the foundation level Equation 1.25 is
applicable.
qnet CNc D(Nq1) subBN
1. 25
For cohesive soils the value of υ is small and the term ɣsubBNɣ is of little account.
Consequently the bearing capacity is not affected by the ground water variation below the
foundation level. For sandy soils the term CNc is zero and the term 0.5γsubBNγ is about half
0.5γBNγ. The effect of the groundwater is significant.
Water table above the foundation level
For this case the net ultimate bearing capacity is given by Equation 1.26. It is seen both
cohesive and cohesionless soils are affected by the water table rising above the foundation
levels
qnet CNc p'o (Nq1) subBN
1. 26
Where p’o is the effective overburden above the foundation level.
1.2.4 Allowable bearing pressure
In design, the settlement due to the safe net bearing pressure is computed. If the resulting
settlement is not acceptable then the pressure used in the determination of the settlement is
reduced. At the point when the settlement is acceptable then the pressure obtained is the
allowable bearing capacity of the soil.
In design the ultimate loads are obtained from structural analysis. The ultimate load
is converted into the service load. The gross load is the structural load above the ground floor
plus the overburden. The net load at the foundation level is the load at the ground floor in
addition to the weight of the foundation less any soil which has been replaced. For practical
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Shallow foundations
considerations it is therefore not necessary to consider the weight of the foundation below the
ground level (Figure 1.9)
P
Gross load =P + overburden
Pnet = P + foundation load – replaced
soil
=P
Figure 1. 9 Net load applied at the foundation level
1.2.5 Field methods for the determination of bearing capacity of soils
Plate bearing test
The test is particularly suited for the design of foundations or footings where it is considered
that the mass characteristics would differ from the laboratory tests and where the precise
values of settlement are required. The plate load test covers the determination of vertical
deformation and strength characteristics of soil insitu. From the data recorded the allowable
bearing capacity of the soil is estimated.
In the test an excavation is made to the expected foundation level. The plate usually
300 to 750 mm square should be rigid and flat. It is loaded by means of kentledge. The
kentledge can be any form of dead load including water, concrete blocks etc or tension piles.
The loading procedure can be either constant rate of loading or incremental loading
procedure as described below:Constant rate of penetration
This test is best suited to undrained conditions. In the test the load is applied in a controlled
manner to enable a continuous and uniform rate of penetration. The load is continued until a
penetration of 15% of the plate width is achieved. The ultimate load is considered to be the
load corresponding to the 15% of the plate width penetration.
Incremental load test
This test is best suited to drained conditions. In the test the load an estimate of the maximum
load is calculated. Five equally spaced increments are then selected. The load at each
increment is recorded together with the corresponding settlement. A load is maintained until
the penetration has ceased or when the primary consolidation is complete. The ultimate load
is considered to be the load corresponding to the 15% of the plate width penetration as in the
case of the constant rate of penetration test.
Plate bearing capacity test results
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Shallow foundations
The plate bearing test results are best reported in graphical way as shown on Figure 1.10.
The weak soft clays and loose sands will reach the ultimate bearing capacity in the region of
100-200 kN/m2 while the stiffer clays and dense sands and gravels will continue increasing
in the bearing pressure with increasing settlement.
Bearing Pressure (kN/m2)
0
200
400
600
800
1000
0
Settlement (mm)
-2.5
-5
-7.5
-10
Stiff clay, dense sand or gravel
Soft clay or loose sand
-12.5
-15
Figure 1.10 Typical plate loading test results
Estimation of allowable bearing pressure from plate bearing test results
The test is reliable only if the stratum being tested is reasonably uniform over the significant
depth of the full scale foundation. A weak stratum below the significant depth of the plate
but within significant depth of the foundation would have no influence over the plate test
results but will have a significant effect over the performance of the foundation (Figure 1.11).
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Shallow foundations
B
b
1.5b
1.5B
Weak stratum
Figure 1.11 Influence of weak stratum
Settlement of the stratum increases with increasing loaded area and the main
problem is in the extrapolation of the test results to full scale scenario. Ideally the plate test
should be carried out using plates of different sizes and at different depths. However, this is
usually not economical.
Notwithstanding this shortcomings the following procedure was been proposed by
Terzaghi and Peck (1948) and can be used as a guide to use of plate bearing test results. The
settlement of a square footing kept at a constant pressure increases as the footing size
increases. The relationship is shown on Equation 1.27 relates the settlement of the test plate
of 300 mm square and that of a square foundation of width B.
S S1 *(2B/(B0.3))2
1. 27
Where S1 = settlement of the loaded area under a 305mm plate for a given pressure
intensity p
S= the settlement of a square foundation of width B in metres under pressure p
In order to use the plate bearing results the maximum allowable settlement is determined. A
value of 25mm is generally accepted as an allowable settlement. S is then equated to 25 and
a numerical value of B is inserted in the formula to enable the determination of the S1. From
the relationship of p and s1 the value of p corresponding to the calculated value of S1 is the
allowable bearing pressure subject to any adjustments certain to the ground water conditions.
Standard penetration test
The test covers the determination of the resistance of soils particularly sand and loose to
medium loose gravels at the base of a borehole to the penetration of a split barrel sampler
when dynamically driven in a standard manner. In addition to the determination of resistance
the split sampler is used to obtain disturbed samples for determination of remolded properties
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Shallow foundations
namely particle size analysis and Atterberg limits when the sample has some degree of
plasticity. When used in gravels the sampler is replaced with a 60 o cone which does not
sample the soil.
Figure 1.12 shows the main features of standard penetration test equipment. The
drive shoe and the sampler consist of 51 mm external diameter and 35 mm internal diameter.
It is 450-600 mm long. This is connected to a drive assembly at the bottom of the boring
rods. A pick and release mechanism which ensures a free fall of a hammer weigh 65
kilograms through of 760 mm + or – 20 mm is used to drive the sampler or the cone in the
case of the gravelly strata
Figure 1.12 Standard penetration equipment.
The procedure requires that the borehole is cleaned carefully to ensure that disturbed soil at
the bottom of the borehole is removed. When boring below the ground water table it is
prudent to maintain the water in the borehole at the same level or higher than the general
ground water. A hydraulic balance is needed to avoid the risk of boiling of the strata at
bottom occasioned by a high hydraulic gradient.
The sampler and the hammer are lowered to the bottom of the borehole. If after
touching the bottom the sampler penetration exceeds 450 mm on its own weight and that of
the hammer, the SPT value also known as N value is recorded as zero. Otherwise after the
initial penetration on own weight the test is driven in two stages known as seating drive and
test drive
The seating drive is the initial 150mm penetration or 25 blows whichever is reached
first. The test drive is the further penetration of 300mm or 50 blows which ever is reached
first. The number of blows for the 300 mm penetration is recorded as the SPT value ‘N’. If
300 mm penetration can not be reached in 50 blows the test drive is terminated. In this case a
hard stratum has been encountered and further driving results in damage of the split sampler.
It is usual to record the blows for every 75 mm penetration. If the test drive is terminated the
penetration corresponding to 25 and 50 blows is recorded.
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Shallow foundations
Interpretation of Standard Penetration Test Results
The pore water pressure generated by the hammer during testing affects the value of N.
When the test is carried out below the water table in fine sand or fine silt the resistance
increases as a result increased pore water pressure which does not dissipate immediately. If
the measured N if greater than 15 a correction as shown on Equation 1.28 is performed.
TrueN151/ 2(N 15)
1. 28
The relative density of a soil affects the N values. Terzaghi and Peck (1948) evolved a
qualitative relationship between the relative density and the standard penetration N values.
Gibbs and Hortz put values of relative density. Table 1.3 shows the two relationships
Table 1. 3 Relationship of N values and the relative density of sands
N value
0-4
4-10
10-30
30-50
50+
Terzaghi and Peck
Very loose
Loose
Medium
Dense
Very Dense
Gibbs and Hortz
0-15%
15-35%
35-65%
65-85%
85-100%
The effective stress at the level of the test also affects the penetration of the SPT split barrel
sampler. This effect can be related to the effective overburden at the level of the testing.
Craig (1986) has summarized the correction of the overburden into Equation 1.29.
N' CN N
1. 29
Where N’=the corrected value of SPT
N=the measured value of the SPT or the true N in the case of the saturated loose
sands and silts
CN=Overburden factor
The relationship of CN and effective overburden is shown on Figure 1.12
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Effective Overburden (kN/m2)
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Shallow foundations
500
400
300
200
100
0
0
0.5
1
1.5
2
Correction factor CN
Figure 1.12 Estimation of N’ from measured values of N (Craig 1986)
Standard penetration resistance increases with increasing particle size, increasing overconsolidation ratio and increasing angle of internal friction of the soil. A correlation between
shear strength parameter and N, and effective overburden is shown on Figure 1. It provides
rough estimate of value of and should not be used for very shallow foundations.
SPT - N
50
40
φ=25
30
φ=30
φ=35
20
φ=40
φ=45
10
φ=50
0
0
50
100
150
200
Effective overburden (kN/m2)
250
Figure 1.13 Correlation between shear strength parameter φ and N and effective
overburden
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Shallow foundations
Estimation of allowable bearing pressure from standard penetration tests
In 1948 Terzaghi and Peck presented a chart as shown on Figure 1.14 for the estimation of
allowable bearing capacity while limiting the settlement to 25mm and differential settlement
to 75% of the maximum settlement. The procedure involves determination of the average
value of N’ from all the boreholes at the foundation level. The allowable bearing capacity for
the widest foundation is determined and then applied to all the foundations. Terzaghi and
Peck based his chart on foundations on unsaturated soils when the water table is at lower than
1.0B below the foundation. Thus when the water table is at 1.0B the reduction of the
allowable bearing capacity is zero. The reduction increases linearly as the water rises. When
the water table is at the ground level the reduction is 50%. Thus the provisional value of
allowable bearing capacity obtained from Figure 1.14 should be reduced by the factor Cw
shown on Equation 1.30
Cw 0.50.5Dw /(D B)
1. 30
Where Dw= depth of the water table below the ground level and D
D =the depth of the foundation
B = the width of the foundation
Figure 1. 14 Relationship between N and allowable bearing pressure
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Shallow foundations
Static cone penetration test
The test apparatus consists of a 60o cone as shown on Figure 1. The cone is subjected to
continuous penetration in the soil the rate of 15-20 mm per second. The tip has electrical
sensors for continuous recording of the resistance and penetration as shown on Figure 1. On
the more sophisticated cones the friction along the cone can be measured. In addition the
water pore pressure can also be measured. At every penetration the resistance is measured as
load/cone area and is plotted against penetration
Figure 1.15 Static cone penetration apparatus
Resistance = load/end area = qc (kN/m2)
0
200
400
600
0
Penetration (mm)
-2.5
-5
-7.5
-10
-12.5
-15
Figure 1. 16 Static cone penetration test results
University of Nairobi –FCE 511 Geotechnical Engineering IV
800
1000
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Shallow foundations
From the data the value of Nγ used in Terzaghi Equation for the Ultimate Bearing Capacity
can be estimated from Equation 1.31 the value of internal angle of friction can be obtained
from Figure 1.5 which then enables the determination Nq and the ultimate bearing capacity.
Other empirical values of qa can be obtained from equations 1.32 through 1.33
N qc /80
qa qc / 30 for B< 1.2m
qa qc /50*((B0.3)/ B)2 for B> 1.2m
1. 31
1. 32
1. 33
Allowable bearing capacity on rock stratum
The bearing capacity of rock is the highest that an engineer can expect to get. In some cases
the intact rock has unconfined compressive strength larger than the strength of the concrete
which goes to the making of the foundation. In this case it is the structural design of the
materials rather than the strength of the rock control the foundation design.
For ordinary structures when site investigation is performed by boring, bedrock need
be proved to a depth of three meters to discount the possibility of isolated boulders (Craig,
1987). When un-weathered rock has been reached in foundation construction, the allowable
bearing pressure is based on the inherent strength or the parent rock. The influence of joints,
discontinuities and shear zones is to reduce the allowable bearing capacity. The rock quality
designation (RQD) defined as the ratio of the total length of core of full diameter and length
greater than 100mm or greater to the length of the core run measures the extent of defects and
has been used in the determination of the allowable bearing pressure as shown on Table 1.
Table 1. Allowable bearing capacity RQD
RQD
100
90
75
50
25
0
Allowable bearing capacity (kN/m2)
29,300
19,500
11,700
6,800
2,900
1,000
Source Peck et al, 1973
Bowles (1982) stated that the settlement is more often the concern than the bearing capacity.
Consequently most effort should be taken in the determination of modulus E and Poisson’s
ratio η so that an estimate of the settlement can be made. Alternatively he suggested that one
should use a large factor of safety on the unconfined compression strength of the intact
fragments obtained from the borings. The factor of safety should depend on the RQD and
typically range between 6and 10.
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Shallow foundations
Tomlinson and Boorman (1986) reported the presumed bearing capacity must not
exceed half of the unconfined compression strength of the intact rock fragments. Ibi (1986)
reported presumed allowable bearing capacity values of various rocks varying from 12,500
kN/m2 for igneous and limestone rocks to as low as 150 kN/m2 for weak un-cemented
mudstones.
Rock strength designations based on the unconfined compressive strengths have
been suggested by BS 5930 (Ibi (1986) and the Canadian Geotechnical Society (Franklin and
Dussealt, 1989) are shown on Tables 1.4 and 1.5 respectively.
Table 1. 4 Rock strength designation by BS 5930
Classification
UCS (kN/m2
x103)
Very
Weak
Under 2
Weak
1.25 to
6
Mod
Weak
5 to
20
Mod
Strong
12.5 to
60
Strong
50 to
200
Very
Extremely
strong
strong
100 to Over 200
200
Source – Tomlinson and Boorman (1986)
Table 1. 5 Rock strength designation by Canadian Geotechnical Association
Classification
UCS (kN/m2
x103)
Extremely
Weak
Under
2
Very
Weak
2 to
6
Weak
6 to
20
Medium
Strong
20 to
60
Very
strong
100 to
200
Extremely
strong
Over 200
Source: Franklin and Dussealt (1989)
1.2.6 Presumed bearing capacity of soils and rocks
It is common to use presumed bearing capacity of soils and rocks. The values used have been
derived after many years of testing and performance monitoring of existing structures. These
values are usually conservative do not consider the overburden above the foundation level.
They can be used as preliminary values for the very large structures where an accurate
bearing capacity at the foundation level is needed. In the case of smaller structures these
valued can be considered as final. Table 1.6 shows the presumed bearing capacity of soils as
suggested by BS8004 (1986), while Table 1.7 shows the presumed bearing capacity values
used in Kenya. It is to be noted that difficult soils such as expansive soils, loose sands and
silts and made up ground should be investigated all the time.
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Shallow foundations
Table 1. 6 Presumed allowable bearing vales (BS 8004: 1986)
Category Types of soils and rocks
Rocks
Non
cohesive
soils
Cohesive
soils
Strong igneous and gneissic rocks in sound
Strong limestone and strong sandstones
Schists and slates
Strong shales, mudstones and siltstones
Dense gravel, or dense sand and gravel
Medium dense gravel or medium dense sand
and gravel
Loose gravel or loose sand and gravel
Compact sand
Medium dense sand
Loose sand
Very stiff boulder clay and hard clays
Stiff clays
Firm clays
Soft clays and silts
Very soft clays and expansive clays and silts
Peat, organic soils, made up ground and fill areas
Value
( kN/m2)
10000
4000
3000
>600
<200-600
<200
>300
100-300
<100
300-600
150-300
75-150
<75
Remarks
The foundations
should be taken to
un-weathered rock
The
foundation
width to be not
less than 1m and
water level not
less than below
the
foundation
level
Soils susceptible
to
long-term
consolidation and
settlement
Not applicable
Not applicable
Table 1. 7 Presumed allowable bearing values in Kenya
Soil and rock
Red coffee soil (Red silty clay)
Medium dense sand
Loose gravel (Murram)
Dense gravel
Compact gravel and weathered rock
Un-weathered rock
Expansive soils, loose sands and silts
1.3
Value (kN/m2)
80-120
100-300
100-150
200-400
350-600
>600
Not Applicable
Proportioning of shallow foundations
1.3.1 Contact pressure distribution
This is the distribution of the pressure below the base of the foundation and the ground. The
pattern of the distribution varies according to the stiffness of the foundation. The stiffness
may be described as yielding (elastic), rigid or flexible
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Shallow foundations
Yielding foundation
The stiffness of such foundation is zero. Here the contact pressure distribution has the same
variation as that of the load. Because of its zero stiffness there will be no moments induced
in the footing. Such a condition exists in fresh concrete before it sets. It has no practical
significance.
Rigid foundations
Contrary to the yielding foundation the rigid foundation has infinity rigidity. They are so
rigid that they do not deflect. Most of the foundations considered in practice are rigid
foundations. The analysis is simple and leads to economical design of the footings.
Flexible foundations
The stiffness of such foundations lies between rigid and the yielding foundations. The
foundations in this category deflect to a certain degree depending on the magnitude of their
stiffness. The analysis of such foundations is complicated but leads to an economical design.
However this is not usually done in practice and is not considered in these notes.
1.3.1 Proportioning the foundations
The proportioning of the foundations is usually the final step in the design of a structure. The
type of foundation, sizes and the level of the foundation depend on the result of the site
investigation. Usually partial factors would have been used in the design of the columns.
However unfactored loads would be used in the proportioning of the foundations. The
factored loads are however required in the determination of the foundation depths and design
of the foundation in accordance with BS 8110 (1997). The general procedure for the design
of the foundations follows the following steps
a)
Evaluate the allowable bearing pressure in a site investigation exercise
b)
Examine the existing and future levels around the structure and take into account the
ground bearing strata and the ground water level to determine the final depth of the
foundation
c)
Calculate the loads and the moments if any on the individual footings with partial
safety factors on the structural loads.
d)
Recalculate the loads and the moments on the individual columns and the walls
without partial factors of safety. In many cases it is sufficiently accurate to divide the
factored loads and moments with 1.45.
e)
Calculate the plan area of the foundation using unfactored loads
The plan area of the foundations is determined assuming that all the forces are
transmitted to the soils without exceeding the allowable bearing pressure. The distribution of
the pressure is assumed to be planar. In no case should the extreme pressure be less than
zero. All parts of the foundation in contact with the soil should be included in the assessment
of the contact pressure. Subsequently the designer carries out the structural design of the
foundations. Typical foundations are now discussed
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Shallow foundations
Strip and rectangular footings
A strip footing is significantly greater in length than in width. This type of foundation is used
to support walls and closely spaced columns. When and individual column is supported by a
footing then this foundation is referred to as a pad footing. When two or more columns are
supported by one footing, this is referred to as a combined footing.
Axially loaded strip and rectangular foundations
The contact pressure of these foundations is considered as uniform when loaded axially. The
pressure under the foundations should not exceed the allowable bearing pressure of the
supporting soil. Figure 1.17 shows the pressure distribution of such foundations.
a) Pad foundation
b) Strip foundation
d) Pressure distribution
c) Combined foundation
Figure 1.17 Pressure distribution below individual and strip foundations under axial
load
Eccentrically loaded rectangular foundations
When foundations are subjected to axial and moments at their foundations the soil pressure
resultant does not coincide with the centroid of the footing. The resulting pressure is a
combination of the compression and the moment stresses. While the columns can in almost
all cases resist the moments it is doubtful that the spread footing can sustain an applied
column moment. The base usually will rotate and induce more moment at the far end of the
column.
In conventional analysis the contact pressure distribution under eccentrically loaded
rectangular foundations (Figure 1.) are derived from the common flexural formula. The
general formula for the estimation of the pressure when there is eccentricity in the y and x
axis is given in Equation 1.34.
(x,y) PAMy Iy *xMx Ix *y
1. 34
Where
σ(x,y) = contact pressure at any given point (x, y)
P = the vertical load
x,y = coordinate of the point at which the contact pressure is calculated
My and Mx = the moment about y and x axis respectfully
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Shallow foundations
Ix and Iy = moment of inertia of the footing area about the x and y axis respectively
=L*B3/12 and BL3/12 respectively.
y
L
Mx and My
P
B
y ex
Mx
ey
x
ex
ex
My
Figure 1. 18 Rectangular foundation eccentrically loaded in two axis
When Equation 1.34 results in negative values in some areas, this means that the foundation
soil is taking tension. It is then necessary to change the dimensions to have only compression
pressure at the base. This is difficult and requires trial and error approach for solution of
maximum and minimum pressures. It is prudent to place the foundation such that that there
is only eccentricity in one axis direction as explained below.
Eccentrically loaded rectangular foundations in one axis
In design it is common to determine the magnitude of the contact pressure at the edges.
Equation 1.34 reduces to equation 1.35 shown below and Figure 1.19 shows the pressure
distribution.
q PBL(16e L)
1. 35
P
P
P
M
L
M
M
e LL
L
L
b) e=l/6
c) e>l/6
e
a) e<l/6
Figure 1.19 Soil pressures below footing
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Shallow foundations
When the eccentricity inside mid-third of the base (Figure 1.19a,e<l/6) the computed
minimum pressure is positive soil pressure and the computed maximum pressure should not
exceed the allowable bearing pressure. At e=l/6 Figure 1.19b the minimum soil pressure q=0
and the footing is fully effective in bearing. This limit of eccentricity means that as long as
the eccentricity is less than l/6 also described as falling within the mid-third of the foundation
the entire footing is effective. When the eccentricity is large (Figure 1.19c) and e>l/6 the
computed minimum pressure is negative soil pressure. This is an indication of a tensile stress
between the soil and footing. This in not feasible and the soil pressure has to be evaluated
neglecting any soil tension. The eccentricity is said to be outside mid-third.
For eccentricity outside middle third with respect one axis the maximum soil pressure
redistributes itself since the base cannot take negative pressure. The distribution of pressure
is triangular and is shown on Figure 1.20. The equations applicable in this case can be
derived as follows:P
M
B
L
L’
L’/3
e=M/P
Figure 1. 20 Eccentrically loaded rectangular out of middle third
L' L e
3 2
and
P q (BL')
2
Solving the two equations to obtain the maximum soil pressure q, Equation 1. is obtained
q
2*P
3B(l / 2e)
1.36
Rectangular combined footings
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Shallow foundations
It may not be possible to place columns at the centre of spread footings if they are near the
property line, near mechanical equipment or irregularly spaced columns. Columns located
off center will result in a non uniform soil pressure. In order to avoid the non uniform soil
pressure, an alternative is to enlarge the footing and place one or more of the columns in the
same footing to enable the center of gravity of the columns loads to coincide with the center
of the footing (Figure 1. . The assumption here is that the footing is rigid. The column loads
are taken as point loads and distributed into the footing. The footings are statically
determinate for any number of columns. The column loads are known and the resulting
pressure is shown in equation 1.37
q P/ A
1. 37
P1
P2
variable
S
Figure 1. 21 Combined rectangular footing
Trapezoidal shaped footings
A trapezoidal shaped footing is required when a combined rectangular footing will not result
in uniform pressure. This is usually so when the space between the combined footings is
constricted as shown on Figure 1.22.
b
a
X’
L
Figure 1. 22 Trapezoidal footing
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Shallow foundations
From Figure 1.22 the position of the centre of area of the footing is x’. The centre of the area
is to coincide with the center of gravity of the loads from the two or more columns being
supported by the trapezoidal footing. The position of the base cannot be extended beyond the
length dimension L. L is therefore a known dimension. The value of the area of the
foundation is obtained from the allowable bearing pressure and the total column loads (
A P/ qa ). . The area of the base is shown in Equation 1.38 and the position of the centre
of the area is shown in Equation 1.39. The solution to the two equations leads to unique
values of a and b representing the dimensions of the trapezoidal footing.
A a b L
2
x1 L * 2a b
3 a b
1. 38
1. 39
From Equation 1.39 and Figure 1.22 it can be seen that the solution for a=0 is a triangular
footing and for a=b it is a rectangle. The solution for a trapezoid footing exists only for
Lx1 L
3 2
Strap or cantilever footings
A strap footing is designed to connect an eccentrically loaded column to an interior column
as shown on Figure 1.23. The strap is used to transmit the moment caused by eccentricity to
the interior column footing so that a uniform soil pressure applied to both footings. The strap
serves the same purpose as the interior portion of combined footing and is used in lieu of
combined rectangular or trapezoidal footing. Equations 1.40 through 1.43 are used to
proportion the footing dimensions. The value of eccentricity e is chosen arbitrary by the
designer. Unique solution of the strap footing is not always possible
R1 *S1 P1S
R1 P1 S1
S
R2 P1 P2 R1
L1 / 2 e x
R1 B1 *L1 *qa
1. 40
1. 41
1. 42
and
R2 B2 *L2 *qa
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1. 43
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S
P1
x
Shallow foundations
e
P2
L2
L1/2
S1
R1
R2
Figure 1. 23 Typical strap footing
Three basic considerations for strap footing design are:a)
b)
c)
The strap must be rigid (Istrap/Ifooting>2. This rigidity is necessary to avoid rotation of
the exterior footing.
The footing should be proportioned to approximately the same soil pressures and
avoidance of large differential settlements
The strap should be out of contact with the soil so that there are no soil reactions and
is weightless
A strap footing is to be considered only as a last option when other options would not work.
The extra labor involved in the forming of the deep beam and accompanying costs make it
only an attractive alternative when other options have been exhausted.
Raft foundations
A raft foundation is a large concrete slab used as a foundation of a several columns in several
lines. It may encompass the entire foundation area or only a portion. Raft foundations are
generally used to support storage tanks, several pieces of industrial equipment or high rise
buildings. Figure 1.24 shows some typical raft foundations
A raft foundation is used where the supporting soil has a low bearing capacity.
Traditionally the raft is adopted when pad and structural wall foundations cover over half the
area enclosed by the columns and the structural walls. However this should be evaluated on a
case by case basis since the raft foundations end up with negative moments and top and
bottom reinforcement. This arrangement could end up being more expensive than closely
spaced pads which require only bottom reinforcement.
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Shallow foundations
(a
(b
(c
)
)
)
(a) Flat slab; (b) Thickened under columns or beam slab (c) Basement walls as part of the raft or
cellular construction
Figure 1. 24 Common types of raft foundations
The advantages of the raft foundations over the other foundations include:a)
b)
c)
The effect of combining the column bases is increase in the bearing capacity of the
foundation. This is because the bearing capacity increases with the breadth of the
base.
The raft foundations bridge over the weak spots
They reduce settlement and are particularly suitable for structures sensitive to
settlement.
Raft foundations are usually designed as infinitely rigid in comparison to the supporting soil.
This assumption simplifies the pressure under the raft to a linearly distributed contact
pressure. The centroid of the contact pressure coincides with the line of action of the
resultant force of all the loads acting on the raft. Figure 1.25 shows the pressure distribution
and the resultant of the vertical loads.
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Shallow foundations
Resultant of column and wall loads
σmin
σmax
Resultant of soil pressure
Figure 1. 25 Linear pressure distribution below a rigid raft
A raft foundation is considered as rigid if the column spacing is less than 1.75/λ. λ is given by
Equation 1.44
Ks *b
4*Ec *I
1/ 4
1. 44
Where Ks = coefficient of sub-grade reaction
B = width of strip of the raft between centers of adjacent bays
Ec = modulus of elasticity of concrete
I = the moment of inertia of the strip of concrete
λ. = characteristic coefficient
Bowles (1982) suggests that the coefficient of subgrade reaction be estimated from Equation
1.45.
Ks 40*F*qa
1. 45
Where F = the factor of safety applied to the ultimate bearing capacity
qa = the allowable bearing capacity
Equation 1.44 is applicable when the column loads do not vary in magnitude by more than
20%. The column loads should also be uniformly spaced. The design of the raft follows the
following basic steps
a)
b)
c)
Compute the maximum column and wall loads
Determine the line of action of the resultant of all the loads
Determine the contact pressure distribution using Equation 1.46. Figure 1.26 shows
the arrangement of the columns and the eccentricities with respect to x and y axis.
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Shallow foundations
P*ey * y P*ex *x
A
Ix
Iy
(x,y) P
1. 46
Where ∑P=total loads on the raft
A = Total area of the raft
x, y =Coordinates of any point on the x and y axis passing through the centroid of
the raft
Ix and I y = moment of inertia of the area of the raft with respect to the x and y axis
respectively
ex and ey = the eccentricities of the resultant force in the x and y direction
It is conventional to obtain the pressures at the four corners and then interpolate in between to
enable the determination of moments and shears for the structural design of the raft
y
P1
My
PP2
2
ex
P3
ex
∑P
B
ey
P4
P5
Mx x
ey
P6
P9
P7
P8
L
Figure 1. 26 Raft foundation plan showing column loads
1.3.2 General consideration in the selection of the foundation depth
Once the geometry of the foundation of the foundation has been found, it is necessary to
determine an appropriate depth of the foundation. The following are general considerations
which the designers should take into consideration.
a)
Usually the foundation should be placed below the depth with minimum moisture
variation over the years. This eliminates the shrinkage and collapse effects of the
foundation soil. In this country a depth of between 1.0 and 1.5 metres is usually
sufficient.
University of Nairobi –FCE 511 Geotechnical Engineering IV
- 35b)
c)
d)
e)
f)
Shallow foundations
The foundation should be placed below top soil and below depths with roots of tress.
The roots are potential water paths which weaken the foundations.
The foundations should be sited with due consideration to existing nearby structures.
The exaction of the foundation in the vicinity of the existing structures could lead to
loss of lateral support of the neighboring structures.
Special attention should be taken to foundations supported on expansive soils and
those on loose sandy silts which are likely to be saturated during the lifetime of the
structure.
For water structures viz: - river bridges it is necessary to take extra care to ensure that
scouring of the foundation vicinity does not impair the safety of the foundation. It is
usual to use gabions in areas where scouring is likely to erode the foundations such as
downstream of box culverts and around abutments and pier foundations
It is preferable to place foundations at one level throughout. None the less if it is not
practical to have the foundations at one level, the change of level should be at one
plane. Sloping foundation levels should be completely avoided even if they are on
rock. There is a risk of the foundation sliding.
1.3.3 Foundations for common buildings
This section deals with foundations for ordinary common buildings. These are single and
double storied buildings with structural walls as the main form of support. The spans should
generally not be bigger than six metres. The buildings are generally on good bearing soils.
The bearing soils include red coffee soils, gravelly soils and firm sandy, gravelly clays. The
footing for these common buildings is shown on Figure 1.27. The 600 mm width is a
practical width which allows masons to maneuver into the trench.
200-150 mm thick
masonry wall
100mm slab with BRC no 65 at the top
face
200-150 DPC
Damp proof membrane
150 mm minimum drop
100-200 mm thick hardcore
dropasountonsd
A minimum of 1000 mm
depth of foundations
600mm wide x 200mm deep
mass concrete foundation
Figure 1. 27 Typical strip footing for an ordinary building
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Shallow foundations
The following are the general considerations in the usage of the standard footing.
a)
b)
c)
d)
1.4
No reinforcement is needed for strips where the load can be distributed through 45o.
The foundations should be excavated and the last 150mm excavation be finalized
when the concreting can be done without further delay. This minimizes the softening
of the foundation
The mass concrete is in mass concrete usually by volume batching to achieve grade
15 concrete. A ratio of 1:3:6 for cement sand and ballast respectively is generally
sufficient.
Reinforced concrete foundations are done for areas with concentrated loads. These are
usually column supports. Grade 25 concrete is the lowest class of concrete allowed in
the new BS 8110, but grade 20 of concrete can be considered.
Foundations on difficult soils
1.4.1 Foundations on expansive clays
Introduction
The problems associated with expansive soils arise as a result of alternate heaving and
shrinkage of the clays. These soils are typically black or grey and are referred to as black
cotton soils in this country. The cycle of expansion and shrinkage is a result of ability of the
clays to take in water and retain it in its clay structure. The water absorption leads to
expansion of the clay and causes strains in the foundation and the structures supported
thereupon. The strains eventually cause the cracks to appear on the walls. The result is
structural safety and aesthetics of the buildings are compromised
The clay minerals include montmorillonite, illite and kaolinite as discussed in FCE
311. The montmorillonite clay mineral is particularly prone to heaving and shrinkage. Soil
having more than 20% of montmorillonite are particularly prone to swelling problems
In addition to visual identification the expansive soils can be identified by assessing
the swell potential of the soils. This is done by conducting an odometer test which measures
the free swell and the swell pressure attained in an odometer when a sample held in an
odometer ring is kept at the same volume as swelling is induced by allowing the sample to
take in water. Some of the Nairobi black cotton soils have been found to have a swell
pressure of up to 350 kN/m2. Chen ( ) has related swell potential to plasticity index as shown
on Table 1.8. The following methods can be applied to mitigate damage control
a)
b)
c)
Moisture control
Soil stabilization
Structural measures
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Shallow foundations
Table 1.8 Relationship of swelling potential and plasticity
Swelling Potential
Low
Medium
High
Very High
Plasticity index (PI)
0-15
10-35
20-55
Over 55
Source (Chen, )
Moisture control
The main course of heave and shrinkage is the fluctuations of moisture under and around the
structures in question. Depending on the topographical, geological and weather conditions
the natural ground water fluctuates during the year. This seasonal fluctuation decreases with
depth. In some areas the depth to the fluation zone is as low as 1.5 meters. In other areas it
will be deeper going down to over three meters. In addition to the ground water fluctuation
the surface water from rains or bust pipes seeps into the foundations and course moisture
migration.
A satisfactory solution to the problem would to devise an economical way of
stabilizing the soil moisture under and around the structure. It does not matter whether the
moisture is maintained high or low in so far as it can be maintained throughout the year. An
effective procedure of achieving this is to provide a water tight apron of approximately one
metre round the building. A subsurface drain one metre round the building is provided with
augur holes provided at every 2 meters. The holes are filled with sand and interconnected at
the top. In effect the augur drain is and the impervious apron ensures that the moisture at the
foundation area remains the same. Figure 1. 28 shows such an arrangement of the drains for
ensures that the moisture content of the foundations remain the same
The subsurface drain is used to intercept the gravity flow, or; perched water of free
water to lower ground. It also arrests capillary moisture water movement. The subsurface
drain should be lend to a positive outlet. In general the ground surface around the building
should be graded so that surface water will flow away from the building foundations all h the
time.
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Shallow foundations
Positive drain to outfall
away from the building
Building
a)
Location of sand drain around a building
Ground floor with double
mesh A142
Masonry
Original ground
walling
level
2 meter wide water
Compacted granular material at
tight apron
high water content
Coarse sand drains
at 2 metre intervals
Expansive soil
b) Sand drain and apron detail
Figure 1. 28 Typical sand drain treatment of a building
Soil stabilization
Soil stabilization consists of one of the following operations
(a)
Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction.
(b)
Compaction control
(c)
Soil replacement
(d)
Chemical stabilization
Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction involves the
flooding of the site under consideration prior to construction. The soil would heave and the
potential danger of cracking is eliminated. Pre-wetting has been used with success when the
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Shallow foundations
active zones are not large. It is very difficult to saturate high plasticity clays. There is danger
that expansion of the clays could continue after the construction has taken place. This
procedure should be considered for stabilizing pavement or canal linings. In only rare cases
should the method be considered for use below ground floor slabs. Its application below
building foundations is risky and questionable.
Compaction control has been used in pavement construction. Expansive clays expand very
little when compacted at low densities and high moisture contents. But will expand
considerably when compacted to high densities at low moisture contents. The approach is to
compact swelling clays at moisture contents slightly above their natural moisture content for
good result. In this method it is not necessary to introduce large amounts of water into the
soil. Dry compaction of expansive soils was done along the Lodwar-Kakuma road.
Soil replacement is the simplest an easiest solution for slabs and footings founded on
expansive soils. The expansive foundation soils are replaced with non-heaving materials.
The method requires the selection of the replacement material and the depth to replacement.
In Nairobi the depth of the expansive black cotton soils is in the region of 1.0 to 1.5 metres.
In this case it has been found desirable to remove the entire expansive soil below buildings
and replace with suitable granular material. When the expansive soil is deeper building slabs
can be constructed above the compacted soil covering the expansive soil but the foundation
of main structure needs further consideration.
This method is particularly useful for the construction of highway pavement in a site
completely overlaid with expansive soils where the alternative to reroute the road is not
viable. In this case it the lower expansive soils are overlaid with the compacted replaced
material to a depth of 1.5 metres.
Chemical stabilization is the process of mixing additives like cement and lime to expansive
soil to alter its chemical structure and in the process retard its potential expansiveness. Lime
reduces the plasticity of the soil and hence its swelling potential. The amounts used range
from two to eight percent by weight. Cement on the other hand reduces the liquid limit,
plasticity and potential volume change. Stabilization has been used mainly in highway and
airport construction.
Structural measures include several methods have been reported in literature such methods
include
(a)
(b)
(c)
Floating foundation
Reinforcement of brick walls
Foundation on piles
Floating foundation concept is a providing a stiffened foundation. This is essentially a slab
on ground foundation with the main supporting beams resting on non-cohesive non heaving
material. The slabs are designed fixed on the beams that assuming a heave pressure of 20
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Shallow foundations
kN/m2. This magnitude is small considering that the swell pressure of the expansive soils
commonly found in Kenya has been estimated at between 300 and 500 kN/m2. Results of
such an approach have been mixed where they have been tried. This method needs further
research.
Reinforcement of brick walls have been tried in South Africa. In this method reinforcement is
placed in brick walls. The reinforcement is placed where cracking usually takes place. This
is typically above and below openings. The structure is made also semi flexible by providing
joints in the brickwork so that when heave takes place the building will conform to the new
ground shape and consequently reduce the bending moment induced in the walls. The joints
are typically 1.5cm.
Foundation on piles is a very successful procedure which ignores the heave by placing the
footing to a sufficient depth (Figure ). The depth of the pile should leave an expansion zone
between the ground and the building to allow the soil to swell without causing detrimental
effect to the building. One way of installing the piles is to provide a pile with bell at the
bottom. The bell or under reamed section should be well below the active zone. The bell is
installed with special equipment and anchors the pile into the ground. The pile can be
installed in an oversize shaft which is subsequently filled with straw saw dust as filler to
eliminate uplifting of the pile by heaving soil. Alternatively the pile could be a straight and
the effect of the uplift calculated using Equation 1.47 The friction below the active zone is
utilized in the calculation of the bearing capacity of the pile.
1. 47
Where
= the total uplift
D = the diameter of the pile
h = the depth of the pile in the active zone
u = the swelling pressure
f = the coefficient of friction between the pile and the soil
f may be taken as 0.15 while the swelling ;pressure varies between 250 and 500 kN/m2
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Shallow foundations
Straight pile
friction
Uplift
zone
zone
Skin
Active
Stable
friction
Uplift
Skin
h1
h
Beam
zone
Stable Active zone
Beam
Under ream
pile
Figure 1. 29 Pile systems for expansive soils
1.4.2 Foundations on loose sands
Foundations on loose sands are particularly difficult due to the likelihood of collapse in the
event of large storms. The storms result in the realignment of the sand particles and
consequent settlement due to repacking of the sand support. This has resulted in large cracks
in buildings which have been placed on this type of foundation soils. The foundation soils
subsequently loose there bearing capacity and the result is settlement of the foundations. The
superstructure has to absorb the settlement usually with resultant cracks of walls and
structural elements.
A real case story is one of the Garissa teachers college whose buildings were placed
on sand strata. The area is generally dry but when the rain comes, it usually very heavy and
comes in large storms. The performance of the three building types of structures adopted at
Garissa teachers college forms a case study whose findings are used to suggest a construction
procedure for foundations and masonry superstructures on loose sands.
The main teaching bungalow consisted of buildings constructed with a ground beam
which was framed with columns and a concrete roof slab. The masonry was thus reinforced
at the corners with columns and subsequently bound at he top by a ring beam and at the
bottom with a ground beam. These types of buildings were found to have performed well
several years after construction. This type of construction produced a satisfactory type of
constructed and when the buildings were inspected ten years after construction the structural
frames and the infill masonry walls were performing well.
The second type of buildings consisted of three and four and three storied flats. As in
the case of the previous buildings these types of buildings were found to have performed well
ten years after construction
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Shallow foundations
The third type of the buildings was the staff residential bungalows. These were
constructed with a ground beam and masonry walls. The roof of the buildings was a concrete
slab. However as the rains came and went in there stormy characteristics the residential
houses developed cracks in the walls. The cracks were particularly severe in the external
walls and after about 10 years of service and needed attention (Plate 1.1
Based on the satisfactory behavior of the framed structures it was found prudent to
introduce columns at the masonry wall corners in a repair scheme. Plate … It is therefore
recommended for foundations on loose sands the masonry should be reinforced with columns
at the corners. In addition the foundations should be kept as far as is possible free from
percolating water. In this way the in the event of settlement the frame will be able to absolve
the stressed attributable to additional settlement and reduce the severity of the cracks.
Plate 1.1 Cracks in the walls occasioned by settlement of the foundation
Plate 1.2 Introduction of columns to stiffen the walls
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Shallow foundations
Tutorial examples on chapter one
1) A footing 2.25 m square is located at a depth of 1.5m. The strength parameters are c’= 0
and υ’ = 38o. Determine the ultimate bearing capacity
a)
b)
If the water is is well below the foundation level.
If the water table is at the surface.
Given that the unit weight of sand above the water is 18 kN/m3. The saturated
unit weight of soil is 20kN/m3.
Ans – A 2,408kN/m3 B, 1,365 kN/m3
2)
A strip footing is to be designed to carry a load of 800kN/m rum at a depth of
0.7m in gravelly sand. The appropriate shear strength parameters are c’= 0 and υ’
= 40o. determine the width of a footing if the a factor of safety of 3 is specified
assuming that the water level may rise to the foundation level Above the water
table the weight of the gravelly sand is 17 kN/m 3. The saturated unit weight of
strata is 20kN/m3.
Ans – 1.55m
3)
A footing 2m square is located at a depth of 4 m in stiff clay of saturated unit
weight 21kN/m3. The undrained strength of the clay at a depth of 4m is cu=
120kN/m3 and υ’ = 0. For a factor of safety of 3 with respect to shear failure,
what load can be carried by the footing
Ans – 1680kN
4)
You are responsible for the design of a combined footing to support two columns
as shown in the figure below. The vertical dead loads on column A and B are 500
and 1400KN respectively. The design requires that the resultant of the column
loads acts through the centroid of the footing. In addition the dead loads, columns
A and B also can carry vertical live loads of up to 800 and 1200 KN respectively.
The live loads vary with time, and thus may be present some days and absent
other days. In addition the live load on each column is independent of that on the
other column. Check that the design meets all eccentricity requirements if the
worst possible combination of live loads is imposed
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Shallow foundations
5) A column is carrying a load of 1200kN. The column is located 300mm form the
boundary of wall. Calculate the pressure distribution if the column is founded on a
square base of 1500mm x1500mm. is the foundation safe if the allowable bearing
pressure is estimated at 300kN/m2
6) An internal column is carrying a load of 2400kN. It is located 3000mm from the
column described in Question 1 Design:a.
a suitable combined base for the two columns
b.
A suitable strap footing for the two columns
7) Your client acquires the next plot and you are not limited by the boundary wall.
Calculate the safe bearing pressure below the columns described in questions 1 and 2.
Assume a detailed site investigation has established the following strength parameters.
C’ = 10kN/m2, υ’ =20o, γsat = 18 kN/m2, γb= 16 kN/m2,
4
Four columns are carrying a tower. If the columns are on a square grid of
2.5mssquare, calculate the pressure at each of the four column positions if a raft
foundation of 3 mmx3m is designed to carry the foundation loads estimated at
4000kN, 5000kN, 6000kN and 7000kN
University of Nairobi –FCE 511 Geotechnical Engineering IV
Chapter two: Deep
Foundations
Deep foundation can be categorized into three major types. These include
i.
ii.
iii.
Pile foundations
Drilled piers
Caisson foundations.
The ground and structural conditions which require the use of the two types are discussed
under each of the sections dealing with the two types of the foundations.
2.1
Pile foundations
2.1.1 Introduction
Pile foundations are structural members used to transmit surface loads to lower levels in the
soil mass. They are used when soil beneath the level at an appropriate raft or conventional
footing is too weak or too compressible to provide adequate support to the structure load.
The piles have small cross-section area compared to their lengths. The pile materials
generally include timber, steel or concrete. The transfer is by vertical distribution of load
along the pile surface and at the pile end point.
Piles may be used in the following circumstances
a)
To transfer loads to a suitable bearing layer when weak strata is ignored and the load
is transferred to an overlying strong bedrock or compact layer.
b)
To transfer load through the shaft friction when compact layer is very deep and would
be impractical to reach it
c)
To support structures over water where conventional exaction and construction of the
foundation is not possible or very expensive to achieve.
d)
To reduce settlement and in particular differential settlement
e)
Based on cost. It might prove economical to drive piles down the strata and then
build on top of the piles instead of having to excavate deep layers and then construct
ordinary foundations
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- 46f)
g)
h)
Deep Foundations
In structures which have considerable uplift, horizontal and/or inclined forces. This is
especially true for marine and harbor works.
To increase the bearing capacity by vibration and compaction of granular layers of
soil.
In soils where deep excavations would result in damage of existing buildings.
Piles can be distinguished by the function they are intended to perform or by the material and
construction procedures used in their construction. The various types of piles by function are
shown on Figure 2.1. The main function of the piles is to take the loads by end bearing or by
friction or by combination of the two. Other functions exist and two which can be sited here
include tension piles and fender piles. The tension piles take lateral forces in place of
traditional retaining walls while fender piles also referred to as dolphin piles are marine
structures principally for taking horizontal loads from vessels in the docking areas. Section
2.2 is presentation of piles by their material and construction procedures.
Soft soil
Soft soil
Soft soil
Friction
Firm
resistance
strata
Hard
strata
End bearing pile
Friction pile
Combination
Impact from floating
object
Tension resistance
Tension pile
Dolphin or fender pile
Figure 2. 1 Types of piles by function
2.1.2 Classification of Piles by materials and construction
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Deep Foundations
Piles are constructed in a variety of properties of materials, construction methods and
functions. This makes as simple classification difficult. Notwithstanding theses difficulties
they are classified in accordance with the pile materials and method of construction (Figure
2.2). This classification also identifies the pile materials. The principal timber materials are
timber, concrete and steel.
Types of piles
Driven piles
Large displacement
Preformed. Solid
or hollow tubes
closed at the end
and left in position
Cast in place formed
by driving closed
tubular sections
and then filling the
void as the tube is
withdrawn
Solid
Pre-cast concrete or
Timber. Formed to
required lengths as
units with mechanical
a) H and
pipe piles
Bored piles
Small displacement
Replacement
Steel sections
H Piles
Open ended tubes
unless a plug forms
during driving
A void is formed
by excavation.
the void is filled
with concrete
sides may be
Supported or
unsupported
Hollow
Steel or concrete
tubes closed at the
bottom. Filled or
unfilled after driving
b) RC
Precast pile
The supporting may effected permanently
by casing or
Temporarily by casing or drilling mud
(Betonite) or
By soil on a continuous auger
c) Shell
Pile
Figure 2.2 Principal Types of piles
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d) Cast in-situ
tube withdrawn
e) Bored pile
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Deep Foundations
2.1.3 Driven piles
To install prefabricated and some form of cast in place piles it is necessary to displace soil by
driving the piles. The piling is commonly done by means of a hammer. The hammer
operates between guides or leads by use of lifting cranes. The leads are carried by the cranes
such that they can drive vertical or raking piles. The piling assembly may be mounted on
base suitable for operation on land or on a floating pontoon in the case of piling in the sea.
The hammers may be free falling operated by a clutch release mechanism.
Alternatively they are powered by diesel or steam. There are several forms of mechanical
devices and equipment in the market used by piling contractors. In order to reduce the
impact stresses on the hammer and the piles it is normal to strike the pile through a hammer
cushion. The elements of cushion vary but are mainly wood packing in a steel cap or dolly.
The various elements in the cushion not only protect the top of the pile but have a significant
influence on the stress waves developed in the pile during the driving. The rating of a
hammer is based on the gross energy per blow. For a drop hammer the rated energy is the
product of the hammer and the height of fall. The efficiency of the hammer is the defined as
the energy delivered at impact divided by the gross rated energy. Energy having been lost in
the dropping of the hammer to pile. For driving piles to great length the hammers have
energies of between of between 50kNm to over 180kNm.
Piles are installed by impact hammers and driven to a resistance measured by
number of blows required in the final stages of piling. For wood piles the energy would be
limited to about 3 to 4 blows per inch when energy of 15kNm is applied by the hammer. If
the pile is to be driven through heaving strata then, it might be necessary to predrill the
borehole where the pile is to be driven. This eliminates undesirable heaving. Additionally if
the pile is to be driven through dense layers of sand and gravel it is possible to loosen the
hard strata by sending a stream of water jet with specially adapted equipment. The various
types of driven piles are now described.
Timber Piles
Timber piles are made of trunks of timber. The timber should be preserved to prevent decay.
Untreated timber embedded below the ground water table has a long life. If the timber is
exposed to alternating wetting and drying it is subject to decay. These types of piles are not
very common.
Steel Piles
Steel piles (Figure 2.2a) are usually in form of H-Piles and pipe piles. H piles are preferred
where high depth is required while the pipe piles are usually filled with concrete after driving.
In the case of H-Piles the flanges and the web are equal thickness in order to
withstand large impact forces. Steel H piles penetrate the ground more readily than other pile
types because of the relatively small cross-section area. They are subsequently used to reach
stronger bearing stratum at great depth. Steel H piles have also relatively large bearing
capacity of between 500 and 2,000 kN per pile depending on the size of the H section. The
pile H sections are usually 250x250 to 350x350 with varying section thickness.
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Deep Foundations
Pipe piles are of the range of 250mm to 750 mm diameter. The wall thickness is
usually over 2.54mm. In the event that the wall thickness is less than 4.54mm the pile has to
driven with a mandrel. When the thickness of wall is over 2.5mm the pipe acts with any
concrete in carrying the load. Pipe piles are usually driven with the lower end closed with a
plate. In some instances conical driving shoes have been attached. The advantage is not
significant.
Steel piles are subjected to corrosion. The corrosion is minimal when the entire pile
is embedded in natural soil. However, the corrosion can significantly increase in the event of
entrapped oxygen. Zones of water table variation are particularly vulnerable. Severe attacks
are encountered on sea structural sections exposed to high and low water tides where the salt
sprays can significantly cause corrosion. The standard practice is to use piles which have a
factory applied epoxy coating. The most vulnerable sections of the piles should be encased in
concrete.
Hard driving and driving through obstructions causes the piles to twist and bend.
They can easily go out of plumb without the piling team recognizing since the depth is at
depth. Deviations from the vertical of below 10% are usually accepted. A penetration of 2 to
2.5mm per blow should be considered as refusal and further driving would generally cause
deterioration.
Pre-cast Concrete Piles
Pre-cast Concrete Piles (Figure 2.2b) are usually cast in a casting yard and transported to the
construction site. Where hard driving is expected the tip of the pile is fitted with a driving
shoe. They are usually of square or octagonal section. The reinforcement is necessary
within the pile to withstand both handling and driving stresses. It is necessary that the exact
length to be installed be determined accurately. If the required length is underestimated, the
extension can be done only with a lot of difficulties. If the length provided proves to be
longer than needed at the site, the piles have to be cut again with a lot of difficulties.
Pre-stressed concrete piles are used and generally have less reinforcement. The prestressing reduces the incidence of tension cracking during handling and driving. The
difficulties related to the pre-cast concrete piles also apply to the pre-stressed concrete piles
Pre-cast concrete piles have relatively large bearing capacity of between 800 and
2,000 kN per pile. The presence of high concentrations of magnesium or sodium sulphate in
the piled environments causes the piles to deteriorate. The deterioration is in the form of rust
in the reinforcement, cracking and spalling. The best practice is dense concrete of high
quality or the use of pre-stressed piles which are not so much susceptible because tension
cracks are minimized.
Driven cast in place piles
Driven cast in place fall in two categories namely case or uncased type. In the cased type
also known as shell the shell type a corrugated steel or pipe which is driven into the ground.
The driving is terminated when the desired length of the pile has been achieved. The
concrete is poured in the shell and left place. In the shell is then left in place. Figure 2.3
shows the schematic installation of a shell type pile.
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Deep Foundations
(1) RC
shells
threaded
on
mandrel and set in position
(2) Pile driven to the required
set
(3) Mandrel is withdrawn and
top shells above the top of
(1)
the pile are removed. A
cage of reinforcement is
introduced
(2)
`
(3)
(4)
(4) Core concrete is inserted
Figure 2.3 Shell type of pile
In the uncased type a steel tube is driven into the ground and tube is withdrawn upon
concreting. Figure 2.4 shows the schematic installation of a typical driven cast in situ pile
where the casing is withdrawn. The pile illustrated is also known as a Franki pile.
(1)
A gravel pug is compacted at
the lower end of the pile tube
(2) Pile driven to the required
set
(1)
(3) Plug broken and a concrete
plug is formed
(4) Core concrete is inserted
(2)
(3)
(4)
(5)
(5) Tube
is
withdrawn
as
concrete is placed
Figure 2.4 Installation of a Franki pile
Difficulties encountered in the installation of driven piles
The installation of driven piles has difficulties due to various factors incidental to the
installation procedures and to the ground encountered at the sites. These difficulties are
varied but the main ones include:a)
Handling of the preformed sections which could lead to damage of the piles before
installation.
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- 51b)
c)
d)
Deep Foundations
Noise arising from the hammer dropping on to the pile. This can be particularly
undesirable in sites in the busy neighborhoods.
Spoiling of the pile in the driving operations include the spoiling of pile heads and or
pile toes. This usually takes place due to overdriving piles when refusal has been
reached. It is usually sufficient to achieve a penetration of 2-2.5 mm per blow in the
last stages of piling.
Piles of small cross-section especially H piles driven in boulderly strata could easily
alignment. Vertical piles could end up having bent up shapes and hence lose their
carrying capacity.
2.1.4 Bored piles
Bored piles are also known as cast in place concrete piles (Figures 2.2c-e. The borehole is
effected by various methods using piling equipment. The bore is supported by casing or by
drilling mud (bentonite suspension). At the required depth boring is stopped and the hole is
filled with concrete. If required a cage of reinforcement is placed before concreting is done.
With the use of bored piles larger diameter piles have been installed with corresponding high
bearing capacities. They are constructed in diameters ranging from 300mm to as high as
2400mm.
They have been performed to depths of 70 metres and below and can be
constructed vertically or in rakes of up to 1:4. They are thus ideal for many site conditions.
The construction sequence of bored piles depends on the method of construction adopted.
The main construction methods include bored piles with casing support and bored piles with
bentonite support.
Bored piles with casing support
In this type of pile the casing is advanced by a crane and a casing oscillator. The material
below the casing area is excavated and brought up for examination and testing where
necessary. After the depth needed has been achieved the reinforcement cage is inserted
followed by concreting as shown on Figure 2.5
Bored piles with bentonite support
In this type of pile a lead casing is advanced into the soil. The material below the casing area
is excavated and brought up by use of drilling equipment with a bucket which can bail out the
drilled soil. The excavated soil is examined and tested where possible. The drilled hole is
supported by drilling mud After the depth needed has been achieved the reinforcement cage
is inserted
Figure
2.5 followed by concreting as shown on
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Deep Foundations
Install casing
Advance the
Insert
Place concrete with a
Complete
using an
casing and
reinforcement
tremie pipe as casing
pile
oscillator
excavate with grab
cage
is withdrawn
This installation is particularly desirable in gravelly and boulderly conditions
a)
With casing
Install
starter
casing
This
b)
Advance into the
Insert
soil by drilling
reinforcement
and supporting
cage
installation
is
suitable
in
all soils
with bentonite
Place concrete
with a tremie
pipe and recycle
bentonite
Complete pile
With betonite support
Figure 2.5 Installation of a bored pile with drilling mud
Difficulties encountered in the installation of bored piles
The difficulties associated with the installation of bored piles are also varied but the main
ones include:i.
Poor base preparation after the bearing strata has been reached. Loose particles will
have reached the bottom of the bore and will be difficult to detect or remove. The
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ii.
iii.
Deep Foundations
base the pile will consequently have a lower bearing capacity than would have been
expected
Poor concreting control where the pile is being cast under artesian conditions. This
usually results from poor shaft control as the concreting continues. The result is
necking of the concrete and/or washout of various sections of the pile. Under ideal
conditions the concreter under tremie conditions should always be placed inside the
wet concrete.
Vibration and movement of the ground in the vicinity of the pile under construction.
It is to be noted that these difficulties are also present in the driven cast in place piles where
the casing is withdrawn as concreting proceeds
2.1.5 Determination of pile load carrying capacity
Determination of load carrying capacity by soil mechanics
Pile design is preceded by extensive site investigation to establish the geotechnical properties
of the soil where the piles will be installed. The parameters obtained in the investigations are
then used in the estimation of the load carrying capacity of the piles. Piles derive their
capacity from base resistance and from side friction. The ultimate load that can be carried by
a pile is then given by Equation 2.1. The terms are explained in Figure 2.6. The accuracy
of the equation depends on the determination of the parameters used in the determination of
Qb and Qs.
Where
= Ultimate Load carrying capacity of the pile
Ultimate Load carrying capacity of the base of the pile
= Ultimate Load carrying capacity of the pile side friction
2. 1
Where
Ab= Area of the pile at the toe of the pile
qf = Ultimate bearing capacity at the toe of the pile
= Surface area of the pile shaft
= Ultimate shearing resistance of the shaft of the pile generally referred to as the
shaft friction
An appropriate factor of safety is applied to the ultimate load. It is prudent to apply different
values for the base and the side friction. This is primarily because the movement needed to
mobilize the friction resistance is much less than the movement needed to mobilize the base
resistance. Initially as the pile is loaded the load is taken by the side friction and as load is
increased the base takes more load. At failure the proportion of load supported by friction
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Deep Foundations
may actually decrease slightly due to plastic flow of the soil near the base of the pile.
Equation 2.2 shows the allowable load when allowing for a factor of safety of 2 and 3 for side
friction and base resistance respectively.
2.2
Qs
Qb
Figure 2.6 Load distribution of load on a pile
Cohesive soils
Base resistance: The base resistance Qb of piles in cohesive soils is based on the bearing
capacity factor Nc .
2. 3
Where
= bearing capacity factor which is usually taken as 9.0
= undisturbed un-drained shear strength of the soil at the base of the pile
= the cross section area of the pile at the base
In the case of driven piles the clay adjacent to the pile is displaced both laterally and
vertically. Upward movement of the clay results in heave of the ground around the pile and
can cause reduction of the bearing capacity of the pile. The clay in the vicinity of the pile is
completely remolded during driving. Excess pore water pressures are set up during driving.
This pore pressure dissipates in a few months and in any case before significant load is
applied to the pile
In the case of bored pile, the clay area around the pile will be remolded. Additionally
as the water seeps towards the created borehole their softening of the soil in the vicinity of
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the pile. Water can also be absolved from the wet concrete when it comes in contact with the
clay. The upshot of this is and subsequent reduction of the pile bearing capacity.
Side resistance is based on the friction mobilized on the surface of the pile. Equation 2.4 and
2.5 shows the estimation of the side friction
̅̅̅
2. 4
̅̅̅
2. 5
Where
= adhesion factor between the pile and the soil
̅̅̅ = the average undisturbed shear strength of soil adjoining the pile
= the shaft area which contributes to the friction resistance
Most of the load of a pile installed in a clay soil is derived from the shaft friction and the
problem usually revolves accurate determination of the value of α. For soft clays driving of
piles tend to increase strength around the pile. A value of α equal to 1 can be used. It is
however unlikely that the soil will not in the long run return to its original soft status after
some time. In over-consolidated clays the value varies from 0.3 to 0.6 (Smith and Smith,
1998). A value of 0.45 is usually used for design purposes.
An alternative is approach is to express skin friction in terms of effective stress. The
rationale of this approach is that the area of disturbance during pile installation is relatively
small. The excess pore water pressure induced in the installation process dissipates ahead of
the application of load.
̅̅̅
́
2. 6
Where
Ks = the average coefficient of earth pressure and
̅̅̅ = the average effective overburden pressure adjacent to the pile shaft
́ = the angle of internal friction of the remolded clay. The cohesion intercept of
remolded clay in an drained triaxial test being zero.
Cohesionless soils
Base resistance: The ultimate bearing load carried by a pile depends mainly on the relative
density of the sand in which it is driven. The ultimate bearing capacity at the base of the pile
is given by
̀
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Where
= The bearing capacity coefficient.
̀ = The effective overburden pressure at the base of the pile
Value of Nq
It is to be noted that the bearing capacity attributable to Nγ usually ignored in pile design as
the value of B is usually small. The values suggested by Berezantzv et al (1961) are often
used and are shown on Figure
100
Nq
10
25
35
45
φ in Degrees
Figure 2.7 Bearing capacity factors for use in pile design
Source Berezantzv et al 1961
Side friction: Meyerhof (1959) suggested the average value of friction to be estimated from
Equation 2.6. As can be seen from the Equation the value of fs continues to increase as the
effective overburden increase. However field tests have shown that the maximum value of fs
occurs when the embedded length of the pile is between ten and twenty diameters. In practice
a maximum value of 100 kN/m2 of fs is taken.
̅̅̅
2. 7
Where Ks = the average coefficient of earth pressure and
̅̅̅ = the average effective overburden pressure adjacent to the pile shaft
= the angle of internal friction between the soil and the pile.
Typical values of and Ks are given on Table 2.1 after Smith and Smith (1998) are shown on
Table 2.1. The ultimate load that can be carried by the pile is therefore given by Equation
2.7.
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Table 2.1 Typical values of
and Ks
Pile material
Ks
Loose
0.5
1.0
1.5
o
Steel
Concrete
Wood
Deep Foundations
20
0.75υ
0.67 υ
Dense
1
2.0
4.0
Source Smith and Smith (1998)
̀
̅̅̅
2.8
Equation 2.8 shows the allowable load when allowing for a factor of safety of 2 and 3 for
side friction and base resistance respectively.
̀
̅̅̅̅
2.9
Determination of piling parameters from in-situ tests
The above equations pose difficulties with respect to determination of parameters for a
cohesionless soil which is difficult to sample in the field in undisturbed condition for accurate
determination of Nq which depends on the internal angle of friction. The value of the angle of
internal friction between the soil and the pile remains at best an estimate.
Consequently it has been found preferable to use empirical correlations based on the
results of standard penetration and those of the Dutch cone penetration equipment. Meyerhof
(1976) proposed the values given on Table below.
Table 2.2 Pilling parameters from standard penetration tests
Driven piles
Type of soil
Sands and gravels
qb (kN/m2)
Non plastic silts
Bored piles
Any types of soils
fs (kN/m2)
Large diameter - ̅
Average diameter - ̅
Large diameter - ̅
Average diameter - ̅
0.67 ̅
Source Smith and Smith (1998)
Where N = the uncorrected blow count at the base of the pile
̅ = the average uncorrected value of the blows over the embedded length of
the pile
D = is the embedded length of the pile in the bearing stratum
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B = the width or the diameter of the pile.
An alternative to the use of the Standard Penetration tests is to use the Dutch cone test results.
The cone penetration results can be seen in
Figure 2.8. The ultimate base resistance is taken as average value of Cr over a depth of 4d as
shown on
Figure 2.8. The ultimate skin friction can be obtained from Table 2.3.
3d
of the pile
Estimated depth
Depth (m)
Cr (kN/m2)
d
Figure 2.8 Typical results from a Dutch Cone Test
Table 2.3 Skin friction (fs) values from Dutch cone test results
Type of pile
fs kN/m2
̅̅̅
Driven piles in dense sand
Driven piles in loose sand
̅̅̅
Driven piles in non plastic silts
̅̅̅
Where
̅̅̅ is the cone resistance along the embedded length of the pile
The allowable bearing load of the pile as before based on the Dutch Cone Test results is
given by Equation 2.9
2.10
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2.1.6 Determination of load carrying capacity dynamic methods
Determination of load carrying capacity dynamic methods is applicable to driven piles. The
basis of derivation of dynamic formula is that a relationship exists between the pile capacity
and the driving behavior during the last stages of driving. The energy from the hammer to the
pile is transformed into useful energy and can be represented by Equation 2.10 in the last
stages of the pile driving
2. 11
Where
M = the mass of the hammer
g = the acceleration of the hammer
h = the drop the hammer
R = the pile capacity
S = the settlement of the hammer as result of the drop h
In practice the above Equation has been modified to take account of several losses which take
place during the driving process. The main losses of energy occur as a result of sound, heat,
friction, quake, losses associated with elastic behavior of the pile and those associated with
the pile head compression. The net energy is equated to the work done in penetrating the
ground by the pile. Figure 2.9 shows the sequence of the pile driving and the
Wh
h
e fW
efeivWh
h
a) Variation of energy upon falling of hammer on to a driven pile
Permanent +Elastic penetration
(sso + spp) +(sep +ses)
(sso + spp =set =s)
(sep +ses )=c)
(sso+ses)
(sso)
(ses)
b) Penetration of pile upon falling of hammer on to a driven pile
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Figure 2.9 Energy and penetration of a pile during driving
The potential energy of the hammer is Wh. Upon contact with the pile the available energy to
drive the pile into the ground is ef.eiv.Wh, where ef is the efficiency upon falling and eiv is
the efficiency upon impact. The penetration of the pile as shown on Figure 2.9b can be
shown to result in permanent ;penetration attributable to the pile and soil spp aand sso. In
addition there will be elastic penetration sep and ses attributable to the pile and soil
respectively. The work done and the pile resistance equation can now be rewritten as shown
on Equation 2.11.
2. 12
Where R = The ultimate load capacity of the pile
= the overall efficiency factor
Equation 2.10 is known as Hiley formula. In the field the final stages of the pile are
monitored and recorded as can be seen on
Figure 2.10. It is usual to drive the piles to a minimum set of 2.5mm. Harder driving only
goes to damage the toe of the pile and could reduce the pile capacity in the process. Pile
driving formulas should be used in the piles driven in sand and gravel and in any case should
be calibrated with a load test.
Elastic comp = c3
Elastic comp = c2
Elastic comp = c1
set = s1
Figure 2.10 Pile driving trace of the final stages
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set = s2
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2.1.6 Determination of load carrying capacity pile testing
The load test is the most reliable of all the methods used in the determination of load carrying
capacity of a pile. In this method a full scale test is carried out on a working pile. Essentially
the pile is loaded and a plot of load versus settlement is recorded. From the plot the
allowable load is computed by one of the many formulas available from literature. Full scale
piles are then constructed to the same specification as the test pile
The test is conducted by loading the pile with kentledge load or by use of tension piles
(Figure 2.11). In some piling contracts the working piles cannot be used as tension piles for
testing purposes. This is primarily because in the cause of piling test the tension piles are
lifted slightly. This could lead to weakening of the working piles.
Kenteledge
Kentledge
Support
Existing ground level
Jack
Test pile
a)
Load resisted by kentledge
Jack
Existing ground level
Test pile
Tension pile
b)
Tension pile
Load resisted by tension piles
Figure 2.11 Methods of testing piles in the field
If the test pile is a purely test pile ahead of the main installation of the pile the maximum load
to be applied is equal to two and half times the estimated safe carrying capacity of the pile.
It is usual to load the pile to 1.5 times the design allowable pile load when a working pile is
tested for ascertaining the integrity of the piles installed.
Maintained load test
The load is applied by maintaining the load in a series of increments. The increments are
usually equal to 20 to 25percent of the design working load of the pile. The subsequent
increments are carried out when the settlement has reduced to less than 0.25mm per hour.
The load is subsequently withdrawn in the same stages as the loading to trace the unloading
curve.
Constant rate of penetration
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Load
In this method the load is applied by a constant rate of penetration by a jack in order to
maintain a constant penetration rate (Figure 2.11b). it is usual to maintain penetration rates
of 1.5mm per minute and 0.75mm per minute in the case of sands and clays respectively.
Interpretation of test results
The results are plotted on a load settlement curve as shown on Figure 2.12. In the two
procedures ultimate pile load is taken as the load which achieves a settlement equal to 10
percent the diameter of the pile as is seen in test pile a Figure 2.11b. (BS 8004). The ultimate
pile load could also be reached when the shear failure of the pile soil interface or the pile toe
occurs (Figure 2.12b). The allowable pile load is obtained by dividing the ultimate load by
an appropriate factor of safety. The factor of safety usually ranges from 1.3 to 2.0
Load
Settlement
Settlement
Time
a) Maintained load test results
a
Ultimate
load (a)
Load
Ultimate
load (b)
b
Penetration =
0.1 pile diameter
Penetration
b) Constant rate penetration test results
Figure 2.12 Pile test load results
The above failure criterion is applicable to normal size piles. In the case of large diameter
piles on rock the ultimate load depends on the capacity of the concrete. This depends on the
stress in the concrete.
2.1.7 Negative skin friction
Negative skin friction is a phenomenon or which occurs in piles when a force develops
between the pile and the adjoining soil in a direction which increases the load on the pile and
or the pile groups. This phenomenon develops when a compressible layer of clay, silt, or
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mud etc settles on account of consolidation which may be initiated by ground water lowering
or increase in overburden pressure.
As clay layer settles, piles are dragged into the soil by the consolidating soil and the
overburden soil. The direction of the friction is reversed increases the load on the pile. The
friction generated on the perimeter of the pile due to this dragging is carried by the column
instead of assisting in carrying he pile load. The effect is to reduce the carrying capacity of
the pile. This is the phenomenon known as negative skin friction
l-fill
Fill
Compressible clay
l-clay
Length of settling soil=l
Figure 2.13. The negative skin friction may be estimated from Equation 212 for
single piles and Equation2. For group piles
Figure 2.13 Negative skin friction
2.13
For cohesive soils fs is can be approximated to
̅ . while for cohesionless soils fs is
equal to
̅ . Where the value of fs is estimated from triaxial testing for cohesive
soils the fs can be taken as 0.5Cu
Where
= the ultimate force generated by the negative friction
= the shearing resistance of the soil
= length embedded above the bottom of the compressible layer
= the pile diameter
= the coefficient of earth pressure at rest
= angle of shearing resistance in terms of effective stress
̅ = average effective overburden pressure
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2.1.8 Pile groups
In practice piles are designed and constructed to work in groups. In construction of a group a
pile cap is cat on top of the piles. The cap is usually in contact with the soil on top of the
piles. The bearing capacity of the group is an arithmetic sum of the piles and that of the cap.
Banerjee (1975) showed that the pile cap could support up to 60% of the applied load. If the
cap is clear of the ground surface piles in the group are referred to as free standing piles.
Bearing capacity of groups
Except for the large diameter piles of over 700mm diameter the piles are usually designed in
groups of three or more piles under a column. The minimum under a foundation wall would
be two per typical cross-section. Typical arrangement of the piles is given on Figure 2.14. In
general the ultimate load capacity of the pile group is not the sum of the loads of the piles in
the group. The ration of the ultimate load for the group to the sum of the loads carried by
individual piles is the efficiency factor of the group.
3 – Pile
4 – Pile
5 – Pile
12 – Pile
Figure 2.14 Typical arrangement of pile groups
For piles in sand, the group action is complicated by dilatancy and densification
characteristics of the sand. When the spacing of the piles is less than eight times the pile
diameter, group action takes place (Department of Navy, Naval Facilities Engineering
Command, 1982). In dense sand the effect of driving piles is to loosen the sand and hence
the angle of internal friction of the sand in the vicinity of the piles. This results in overall
reduction of the pile bearing capacity. The group efficiency factor is less than one. In loose
sand the effect of driving piles is to increase the density of the sand. The bearing capacity of
the loose sand will therefore be increased. In this case the efficiency factor is more than one.
An efficiency factor of 1.2 is often used. In the case of bored piles in sand the resulting
loosening of sand in the boring operation results in efficiency factors less than 2/3. The
difficulties in the quantification of the design parameters of either loosened or densified sand
strata in piling operations remains a real problem for engineers (Mwea, 1984). Nonetheless
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Deep Foundations
experimental evidence has it that the piles at the centre of a group in sand carry more load
than the piles on the periphery.
For piles in clay the effect of the pile group is to reduce the bearing capacity of the
pile group. This is because the effect of placing piles in a group is to have one large block
taking friction on the sides and base resistance over the block base. The spacing of piles in
clay is of the order of two times the pile diameter to four times the diameter. The efficiency
of the groups range from 0.6 to unity as the pile spacing increases from two diameters to four
diameters. The ultimate load in the case of a pile group is given by Equation 2.13. In the
case where the pile cap rests on the ground the ultimate load should be taken as the less of the
block capacity or the sum of the individual piles on the group.
2. 14
Where
= The width of the group
= Length of the base of the group
= Depth of the group
= Bearing capacity factor of the clay
= The average undrained strength of the undisturbed clay
Whitker (1957) in a series of model tests showed that block failure as a group in clays occurs
when the spacing of the piles is not more than 1.5d apart. General practice is however to
space the piles at between 2 and 3d. In such cases the efficiency of the group is
approximately 0.7.
Settlement of groups
The settlement o a group of piles can be estimated by assuming that the entire load acts at a
depth as an equivalent raft. In clays the raft is assumed to be located at a depth of 2/3 D
where D is the depth of the pile group. The load is at spread of 1:4 from the underside of the
pile cap to allow for friction transfer. After the assumed depth of the raft the load is
distributed at a spread of 1:2 (Error! Reference source not found.a). Immediate settlement
and consolidation settlement can then be estimated for the layers of soil below 2/3D by
application of normal methods.
For groups in sand the equivalent raft is at a depth of 2/3Db from depth 2/3D. The
spread from the perimeter of the piles is 1:4 followed by a spread of 1:2 Error! Reference
source not found.b). The settlement of the underlying sand stratum is then gotten from
application of standard penetration data and or the cone penetration resistance
1:4
D
1:2
2/3D
2/3D
Db 1:4
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Deep Foundations
1:2
Position of equivalent raft
Position of equivalent raft
Clay stratum
Sand stratum
Figure 2.15 Equivalent raft concept for piles
2.2
Drilled piers and Caisson Foundations
2.2.1 Drilled piers
The term drilled pier foundations is used in a number of situations which to refer to deep
foundations which method of construction is fundamentally different from that of piles. A
large shaft performed in soil and then filled with concrete may be termed as a drilled pier.
ACI (1972) refers to all shafts where a person may enter and work as a drilled piers. In this
definition all shafts larger than 750mm diameter can be referred to as drilled piers. Figure
*** shows typical piers used in practice. In general drilled piers are used where the soil has
a low bearing capacity and it is necessary large loads to firmer stratum and the following
conditions preclude the use of smaller piles.
i.
ii.
iii.
Pile vibrations are not acceptable.
Pile members are too small for the loads.
A large bearing end is needed for higher load capacity
Straight pier
Underreamed pier
Pier socketed Into
Rock
2.2.2 Caisson Foundations
The term caisson is also used to refer to box type structures consisting of many cells built in,
concrete or steel or combination of both. They are built wholly or partly at higher ground and
sunk to final position. They are used to transmit large loads through water and soil to firm
strata. They are used in large bridges, shore protection structures. They are generally used
under the following conditions.
i.
The soil contains large boulders which would otherwise obstruct the penetration of
piles and or construction of cast in place piles.
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Deep Foundations
A massive substructure is needed to extend below the river bend to provide resistance
against floating objects and scour.
Foundation is subjected to very large lateral forces.
iii.
Caissons may be divided into three categories
i. Open caissons
ii. Pneumatic caissons
iii. Box caissons or floating caissons
Open caissons
An open caisson essentially consists of a box open at the top and bottom ( Figure 2.16). the
soil is removed from the caisson by grabbing, dredging from inside the caisson. The sinking
of the caisson proceeds by the caissons self weight assisted by cutting edges of the walls.
When the desired level has been reached concrete is poured under onto the base of the
caisson by tremie pipe. In some cases the caisson has been pumped out. But in most of the
cases the caisson has been left in place. The bearing capacity of the soil below is usually
determined by normal bearing equations.
The concrete seal at the bottom is placed as a plug at the bottom of the caisson but
later serves as a permanent base of the caisson. Its thickness can be obtained from the
equations below
For circular caissons
√
For rectangular caissons
√
Where
= thickness of the seal
σo = contact pressure or hydrostatic pressure
R = radius of the caisson in the case of circular caisson
fc = the allowable concrete stress in tension (0.1 to 0.2cube strength)
b= width or the short side of the caisson in the case of a rectangular caisson
l= length or the long side of the caisson in the case of a rectangular caisson
β = coefficient which depends on the l/b ratio
Water level
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Deep Foundations
Ground surface
Cutting edge
Box caisson
Circular open caisson
Figure 2.16 Open Caissons
Pneumatic caissons
Pneumatic caissons provide an airtight enclosure (Figure 2.17). In effect water is prevented
from getting into the enclosure and the workers can excavate and pour concrete under dry
conditions. The reliability of the quality in this case is better in so the mechanical ventilation
is carried out to the strictest of the specifications. Pneumatic caissons are costly and should
be considered only with the following conditions in mind:
i.
ii.
iii.
Premium pay because of associated health hazards
Overall safety requirements are high
Much of the effort is towards making the work environment suitable for the workers
When the excavation has reached the desired stratum the concrete is sent down to the
working chamber carefully to fill any weak points on the exposed strata. After this initial
filling the area is filled except a small portion of the chamber below the roof of the chamber.
This final portion is filled with grout which also fills any spaces which might have been left
behind during the concreting.
The seal design and estimation of the bearing capacity is the same as that of the open
caissons
Compressed air in
working chamber
Figure 2.17 Pneumatic caissons
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Deep Foundations
Box caissons
Open caissons are usually cast on the ground and then towed to the site. They area then
lowered to a prepared ground. They are carefully aligned on place and then made stable by
placement of ballast. The design and construction of box caissons do not bring any new
design requirements. The ground upon which the caisson is being laid needs to have been
exhaustively investigated to ascertain the foundation depth and any likely difficulties likely to
be encountered. After the caisson is in place it may be filled with either sand concrete or
sand. The caisson should be checked against stability as it is floated to the final place of the
intended foundation.
Design of caissons
The caissons will be designed to resist vertical loads including superstructures, own weight
minus buoyancy forces. The lateral forces will typically include forces due to wind,
earthquake, earth and water pressures, and traction from traffic and pressure from current
flow.
The forces acting on a caisson must be estimated as accurately as can be to enable a
safe design. There are many methods adopted by various geotechnical engineers but the for
stability of the caisson the following combination of forces will suffice
i.
ii.
iii.
All forces are resolved into
A single vertical force
Two horizontal forces in the direction across and along the caisson.
It has been found out that analysis of the caisson in a direction transverse to the direction of
the axis is more critical. From Figure ***-* the three equations of static equilibrium are
solved. This are
W = Base reaction + skin friction
Q = Passive pressure created on BF – Passive pressure on DE – Base friction
Q (H+D) = Moment of all the forces
Q
Q
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From structural
analyses
h
D
From geotechnical
analyses
O
From structural
W
analysis
Q
From
geotechnical
analysis
H
A
D
F
D1
O
E
B
γD(Kp-Ka)
C
γD(Kp-Ka)
Qmax =Area ABC-Area FEC
Qmax =1/2 γD2 (Kp - Ka)- ½*2* D (Kp - Ka)*D1
Moments about O:
Qmax (H+D)=1/2 γD2 (Kp-Ka)D*1/3- ½*2* D (Kp-Ka)*D*D1*1/3
Therefore D1 and Qmax can be calculated and necessary adjustments of the caisson are
made depending on values of Kp and Ka
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2.4
Deep Foundations
Examples of Piling Schemes
Sutong bridge in China
Sutong bridge in China (Plate 1), which has a centre span of 1088m, designed in an area of
high winds and likely to be hit with massive earthquakes (Bitener et al, 2007). The
foundation strata presented the designers with particularly difficult task. The soils at the site
consisted of firm to stiff clay extending to 45 metres below the sea bend. This clay strata was
underlain with a medium to very dense coarse sands, silty sands and occasional loam layers
matrix to a to of 250 metres below the sea bed where the basement rock was encountered.
The designed pile groups covered a plan area of 113.8x48.1m. The design consisted
of 2.8 and 2.5 diameter piles. Permanent casings were installed to a depth of 40 metres. The
overall depth of the piles was of the region of 110 metres. The shafts were designed to mainly
be carried by friction since the displacement needed to mobilize the end bearing is two to
three times that needed to mobilize the skin friction The tips of the pile shafts were however
grouted to increase the bearing capacity of the piles. This procedure densifies the soil below
the shaft and any debris left during the drilling operations. The increased the pile capacity
end bearing capacity is of the order of 20%.
Plate 1 of the Sutong Bridge in China (1088 m center span)
The Nyali bridge in Mombasa
This is a pre-stressed concrete bridge founded on seabed which had coral deposits, sand and
clay soils matrix proved to a depth of 100metres below the sea bend. The designers
depended on the skin friction for the centre piers. The design consisted of 2.0metre diameter
shafts drilled down to depth of 50 metres. On plan the piles have a rectangular layout of 3x8
piles per pier.
2.5
Tutorial examples on chapter two
1)
A single pile 0.6 m diameter is bored into sand strata six meters thick overlying a clay
stratum of infinite depth. Detailed investigations have established that N value in the
sand zone increases with depth (n=3Z). The undrained cohesion increases with depth
(Cu = 5+4Z). Assuming the adhesion factor α = 0.35, determine
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Deep Foundations
a)
2)
An equation for the estimation of pile working load if the pile is to terminate
in the sand zone.
b)
An equation for the estimation of the pile working load if the pile is to
terminate in the clay zone.
A precast reinforced concrete pile measured 450mm x450mm. The pile was driven to
a depth of 15 metres to a set of 3mm by a drop hammer of 2.5 tones freely through 1.5
metres. The piling arrangement was changed to have a 4.2 tone hammer falling
through 2 metres. Assuming the same resistance with the new hammer, determine the
set achieved if the following information is also available.
2.5 tone hammer
0.5
4mm
4.5mm
Overall efficiency factor
Elastic compression of pile
Elastic compression of soil
4.2 tone hammer
0.35
4mm
5.0mm
3)
A pile under test has started showing considerable settlement under load of seventy
tones. The pile diameter is 500mm and a length of 8.5metres in stiff clay. Assuming
below the 8.5metres the clay was soft clay and did not contribute to any resistance
evaluate the magnitude of the unit shear along its skin. (Answer 10.5tones per m2).
4)
A 500mm diameter bored pile is to be made in stiff clay to a depth of 20metres. The
un-drained strength of the clay varies with depth as shown in the following table
Depth
4
2
Cu (kN/m ) 78
6
86
8
102
142
132
16
157
20
184
24
212
Determine the maximum load that may be applied to the pile. The following factors
may be taken.
Adhesion factor α = 0.45
Overall factor of safety = 2
Nc for piles is usually taken as = 9
(Answer 1025kN).
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Chapter Three: Introduction to Earth Dams
3.1
Introduction
Advances in geotechnical engineering have enabled design and construction of high dams
impounding large amounts of water. The design and construction follows well documented
procedures gained over the last years from design construction and monitoring of both
successful and unsuccessful projects. The procedures now taken include
i)
Thorough pre-design and preconstruction investigation of the dam foundation
conditions and of the construction materials and design of dams.
ii) Application of engineering skills and techniques to design
iii) Carefully planned and controlled construction
iv) Carefully designed and installed instrumentation and monitoring of the completed
dams
The design and construction of a dam is not complete without accomplishing its intended
purpose and has proved it safe over several cycles of the performance. Carefully designed
and constructional dams are in excess of three hundred meters high. Our own Thika dam
which supplies the Nairobi residents with water rises some sixty three meters above its
foundation.
Failures in dams have been occasioned by improper design, inappropriate
construction methods, including preparation of foundations, placement of the dam
embankment layers, without the necessary controls of compaction control and monitoring.
The design and construction should not be stereotyped on existing dams. Rather each dam
should be unique and dependent on the geology of the available materials. As one embarks
on the design of dams it should the course of dams the causes of failure of dams has been
listed by Singh and Prakash (1985) as shown on Table 3.1
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Introduction to dam design
Table 3.1 Causes earth dam failures
Cause of Failure
Overtopping
Seepage effects (piping, sloughing etc)
Slope slides
Conduit leakage
Damage to slope protection
Miscellaneous
Unknown causes
% Occurrence
30
25
15
13
5
7
5
Type of failure
Hydrological
Geotechnical
Geotechnical
Geotechnical/Structural
Maintenance
General
General
Source: Singh and Prakash (1985)
The design and construction techniques covered in this chapter are applicable to all dams.
However the design and construction of small dams in Kenya is well covered in the manual
prepared by Ministry of Water (1985). Small dams are those whose height does not exceed
15 metres and or its impounded volume does not exceed one million cubic meters (Bureau of
reclamation 1985). The procedures covered in this chapter are inappropriate for the design
and construction of dam materials presenting the followings characteristics
i) Extremely soft, or dispersive or materials with high plasticity
ii) Exceedingly pervious foundations
iii) Exceedingly fractured foundations
These conditions require specialized testing and analysis of the presenting conditions in order
to arrive at an appropriate design
3.2
Selection of type of earth dam
The scope of dams covered is those dams where the major portion of the embankment is
constructed in successful layers compacted in layers. The layers are well bonded into one
another to achieve the necessary requirements of the particular layer. The materials are
borrowed from borrow pits and from the reservoir area of the dam. Earth dams fall into three
categories namely, diaphragm, homogenous and zoned
3.2.1 Diaphragm types
This type of dam is constructed with pervious materials namely sands, gravels and or rock.
An impervious diaphragm is constructed to act as the main barrier to seepage. The
diaphragm is usually made of concrete, or bitumen. Alternatively they are made of thin
compacted earth. In this case the width of the diaphragm at any depth is either less than three
meters or it thickness at any elevation is less than the height above that elevation. Figure ***
shows typical diaphragm type dams
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Introduction to dam design
As with all dams the diaphragm dam should be designed and constructed with care
and precision. All internal diaphragms whether made of rigid materials like concrete or even
compacted earth have potential of cracking caused by differential movement of induced
during consolidation of the dam embankment materials, fluctuating water levels or settling
foundations. Internal concrete diaphragms can not be readily inspected. Earth diaphragms
on the surface require protection with filters, protection against erosion and wave action.
These types of diaphragms are unusually protected by rock fill and rock riprap. The earth
diaphragm is also not readily inspected during routine or emergency inspections. The earth
diaphragms are usually protected from internal erosion by filters usually in the form of
geotextiles.
If most of the material in a diaphragm dam is rock, then this type of dam is referred to
as a rock dam discussed below.
3.2.2 Homogenous types
These types of dams are made up of single kind of material save for the slope protection. The
material in this type of dam must be sufficiently impervious to act as the barrier for the
seepage. Because the impervious materials are inevitably clays which are weak in stability
but good as barrier to the seepage the slopes tend to be rather flatfish. The usual slopes on
the upstream side of the dams are 1:3.5 to 4 while for the down slope slopes need slopes of
1:2.5 to 1:3. Figure *** shows a typical homogenous slope with three flow lines. As can be
seen, seepage inevitably appears on the downstream side at a height of about 1/3 of the height
of the dam.
Rock toes and horizontal blankets are usually used to avoid the seepage breaks on the
down slope side of the dam. Riprap protection is also used on the upstream side to arrest
erosion occasioned by the waves on the upstream side of the dam. Drainage and filter layers
are designed to meet filter requirements. Inclined filters in combination with horizontal filters
built with well graded sand and surrounded by geotextiles have become a normal practice.
Because modification of the homogenous dams has led to successful dams the use of
completely homogenous dams is now not allowed. The homogenous dams are preferred
where other materials of contrasting permeability are unavailable. Alternatively they should
be used where impervious material forming the embankment is abundant and available
principally in the dam area and within the vicinity of the dam.
Zoned types
In this dams, a central core is of impervious material is franked by more pervious materials.
The design of these dams requires that the permeability of dam embankment materials
increases from the core to the outside franking shells. The materials enclose support and
protect an internal impervious core. The upstream sections provide stability during rapid
drawdown. The downstream pervious materials act as drainage to control the line of seepage.
It is usual to place a filter material between the impervious material and the downstream
pervious materials.
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Introduction to dam design
The impervious inner layers are basically clays typically the red coffee soils. The
pervious layers are sands, gravels, cobbles, boulders and rocks. If a variety of soils are
available the type of dam of choice is the zoned dam (Bureau of reclamation 1985). It has
inherent advantages of stability and reduced seepage across the dam wall.
3.2
Design Principles
The dam should be constructed so that a satisfactory performance at minimum cost is
attained. The maintenance costs should also be factored to ensure a facility with the least
maintenance of the upstream, downstream and the apartment structures and the electro
mechanical structures. An earth dam must be stable during all phases of the construction and
the operation of the reservoir. To accomplish this, following criteria must be met:
i)
ii)
iii)
iv)
v)
vi)
vii)
The embankment, foundation, abutments, and reservoir sides must be stable and
should not develop unacceptable deformations during construction or during the usage
of the structure
Sufficient seepage control must be ensured to ensure that excessive piping, instability;
sloughing, material erosion is under control. Additionally the loss should be such that
it dose not impair the intended usage of the facility by excessive loss of water.
The reservoir sides should be stable under all operating conditions to prevent
landslides into the reservoir. It is to be noted that a landslide into the reservoir could
cause large wave to overtop the dam
The embankment must be provided with adequately sized spillway which allows
design flow floods to pass without overtopping the embankment.
Free board allowance should be sufficient to prevent waves from overtopping the
dam.
The dam should be provided with camber which allows settlement of the foundation
and the abutment to take place. This camber is not included in the freeboard
calculations.
The upstream slope must be protected against the wave action while the down slope is
protected against rain erosion and animal grazing
3.3.1 Foundation design
Foundations of dams refer to the dam embankment wall floor and the sides of the
embankment in touch with the original ground of the dam. Foundations are usually not
designed but they require attention to ensure satisfactory performance. The requirement of
the foundation is to be stable under all conditions and to offer sufficient resistance to seepage
to prevent loss of water.
To determine the seepage and stability conditions of foundations the permeability of
the foundations strata in various directions and at various depths need to be determined. In
addition the strength of the strata should be determined by use of appropriated field testing
accompanied by field testing. For small dams however it is normal to use empirical approach
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Introduction to dam design
in the treatment of the foundations. Because the foundations of different materials demand
different treatments the foundations are grouped into three different classes. These classes
can be grouped into
i) Rock foundations
ii) Foundations of coarse grained materials (sand and gravel)
iii) Foundations of fine grained materials (silt and clay)
Rock foundations
Ordinarily the rock foundations do not present any bearing capacity problems. Instead it is
the seepage problems which have to be addressed A thorough site investigation should be
undertaken to establish faults and any areas of excessive weathering which could lead to loss
of water. The procedure would be usually to perform in-situ tests to determine the
permeability of the rock structure. This is undertaken together with a site survey of the
fissures of the rock. If excessive erosive leakage , uplift pressures, high water pressures can
occur though rock crevices, fissures, permeable strata, and/or fault planes, consideration
should be made to grout the foundation.
The foundation grout is basically injection of a sealing material under pressure into
underlying formations through grout holes. Grout ordinarily consists of cement water
mixture in the ration of 10:1 in the case of rich mix to 0.8:1 in the case of a lean mix. Some
additives to the cement water mix is usually done to improve the pumping. The most used
additive is betonite
The injected grout eventually fills the cavities and potential avenues of water.
Grouting is a procedure requiring specialized personnel and equipment to effectively carry
out the operations. In general a centerline curtain of grout of holes spaced at three to six
meters is adequate. Where large zones of fracture occur below the dam wall and in the
immediate upstream of the dam a blanket grout on grid is desirable. The depth of the grouting
is usually in the region of three to ten metres. In most cases a blanket grouting of the
foundation directly below the impervious zone is desirable.
Sand gravel foundations
Generally these foundations have sufficient strength to adequately support the loads induced
by the embankment and the reservoir. Nonetheless exploratory and analysis of the strata
must be carried out as a matter of routine. The main problems of these foundations are under
seepage and subsequent forces exerted by this seepage. These undesirable effects should be
analyzed and mitigated in design and construction.
Foundations on looses sands are suspect and should generally be avoided as the sand
has the potential of collapse under load. These type of foundations should be avoided or
specialized advice sought.
The amount of under seepage should be estimated from values of coefficient of
permeability of the strata. The coefficient of permeability of the strata should be determined
by established methods including pump out tests, tests conducted by observation of boreholes
when pumping is performed in a test borehole or pump in tests as described in FCE 311. The
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Introduction to dam design
magnitude of the seepage forces should also be determined by analyzing the flow net of the
water flow under the dam. This topic has been covered in FCE 411.
The various methods of treatment of the foundations of sands and gravel should aim
at economical control of the under seepage and the control of the subsequent seepage forces
to prevent the undesirable effect of foundation erosion and piping at the exit of the dam.
Excessive treatment of a detention dam might not be necessary while treatment of foundation
of a water supply dam might be prudent. The various treatment techniques are now presented.
i) Cutoff Trenches
Where possible this is usually the treatment of choice. The cutoff should extend down to
bedrock or to other impervious strata. This treatment ensures no future difficulty will be
experienced in piping and or uplift of the dam. A minimum width of the cutoff trench is
shown on Figure 3.1
h
Sand gravel
d
w
Rock
Figure 3.1 cutoff trench
ii)
Partial cutoff
A cutoff that does not go all the way to the foundation, rather it is designed to a proportion of
the depth to the rock or to impermeable layer.. the reduction in area is not proportional to the
reduction in the flow. Thus the reduction cannot be estimated from the flow equation
The action of the partial cut off is similar to that of an obstruction of in a pipe. The reduction
in flow is not proportional to the reduction in the area of the pipe. Experiments have shown
that a 50% cutoff results in 25% reduction of the seepage while an 80% cutoff results in 50%
reduction in the seepage.
iii)
Sheet piling
This is an expensive method of cutting of the seepage through the foundation of an earth
dam. Additionally the seepage continues to pass through the sheet pilling interlocks. It has
been used sometimes in conjunction with the cutoff trenches. The sheet piles cannot be
performed in cobbles and boulders
iv)
Slurry trench
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Introduction to dam design
This is a trench excavated and filled with concrete below the impervious layer. The trench is
kept in position by placing bentonite before concreting to form
v)
Grouting
Various materials have been used to develop grouting procedures to improve the stability and
reduce the permeability of pervious foundations under dam walls. These materials include
a)
b)
c)
Cement – water
Cement – bentonite – water
Bentonite
It is to be noted that grouting is usually an expensive process and it should be allowed after
extensive testing and evaluation.
vi)
Upstream blankets
These are usually made of same material as the impervious core material. In effect the path
of the underseepage is increased and hence the loss of water is reduced.
vii)
Downstream embankment toes
The aim of these blankets
a)
To reduce uplift pressures at the exit of the dam
b)
To readily permit discharge
c)
To prevent piping of the fines
d)
To convey the discharge
Achieved by
a)
Extending the downstream zones
b)
Pervious downstream shells or by use of horizontal blankets
c)
By use of toe drains
3.3.2 Embankment Design
Embankment slopes
The design of and earthfill dam embankment needs combination of many parameters. Many
of these parameters are difficult to determine accurately. These parameters include gradation,
composition and corresponding behavior of the soils under different conditions of saturation
and loading. The stress – strain relationships can be very complex. The result of these
difficulties is that the design of earthfill dam embankment depends largely on successfully
designed, constructed and well performing dams.
Notwithstanding these acknowledged difficulties it is now possible to model out complex
conditions of an earthfill dam embankment. The design of any earthfill dam is preceded by
extensive site investigation to determine the strength and permeability characteristics of the
embankment materials. This enables the design of h the slopes to the embankments to be
checked under the follow conditions.
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Introduction to dam design
i)
Stability during construction and end of construction. In this condition the
embankment has not had the time to have the pore water in the foundations and the
embankments drained. The pore water pressures are highest in the embankment
materials. The strength parameters applicable are the undrained parameters.
ii)
Steady seepage conditions. The core of the dams act as the water barrier of the earth
fill dam. However even the tightest of the clay cores will allow some water
penetration. The rate of penetration will depend on the permeability of the core
material an in due time will reach steady seepage conditions where a phreatic surface
will be developed at the highest level in the embankment. The steady seepage
conditions is critical for the downstream slope. Under these conditions the water has
been impounded the seepage has stabilized through the embankment. The flow net
has been established. All the excess pore water pressures have dissipated. The slopes
of the dam are checked using drained parameters of the foundations and the
embankment materials. The downstream slope is in critical condition during the
steady seepage
iii)
Rapid drawdown conditions. Under these conditions the stabilizing effects of the
water in the reservoir has been removed on the upstream slope. The rapid drawdown
leaves high pore-water pressure in the embankment. The upstream slope is usually in
its weakest state. The upstream slope of the dam is checked using drained parameters
of the foundations and the embankment materials. It is to be noted that a drawdown
of up to 40 meters per day is considered as rapid.
iv)
Stability under severe seismic conditions. The above conditions should be subjected
to acceleration of the embankment occasioned by seismic activities
v)
Protection against erosion. The upstream slope is likely to be subjected to erosion
arising out of the wave action and sloughing as the level of the water fluctuates. This
is mitigated by use of appropriate upstream protection by use of stone riprap. The
downstream slope is subjected to erosion a result of the precipitation and made worse
by grazing in some dams in communities in need of pasture. The usual practice is to
fence off the dam area and to plant grass and appropriate trees.
The stability check is usually to ensure that the shear stresses induced in the embankments
are resisted by the mobilized shear strength. The shear stresses are from the externally
applied loads which include reservoir weight and earthquake forces. Additionally internally
generated forces from the self weight of embankment The shear stresses at the slopes being
checked can be shown on Figure **** below the shear stresses to be resisted is shown on
Equation 3.1
University of Nairobi –FCE 511 Geotechnical Engineering IV
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Introduction to dam design
σ1
σ
τ
σ3
θ
̅
3.1
The external and internal forces produce a compressive stress along the sliding surface. This
mobilizes the shearing strength which resists shearing along the surface being checked. The
shearing strength is given by Equation 3.2
̅
3.2
It is to be noted that while the shear strength is reduced by the increase in the pore water
pressure the shear stress remains the same. This shows the need of understanding and taking
care of the changes in the pore water regime. In practice the design involves the checking of
the slope stability and application of a suitable factor of safety
Compaction
Compaction of earthworks is a key activity to ensure that the envisaged strength and water
tightness is achieved. When the compressibility and loading of the embankment are constant
the more saturated the soil is the higher the likely hood of developing high pore-water
pressures. To minimize the development of high pore-water pressures it is compact the
earthworks just dry of optimum. However for low dams it has been found satisfactory to
compact earthworks at MDD and OMC.. At this moisture content the material is able to
conform to the shape of the foundation and the abutments.
3.3
Inspection of existing dams
General appearance
i)
Sagging crest
ii)
Slope failures
iii)
Wet patches
iv)
Slope protection
v)
Soil erosion – gullies
vi)
Loss of riprap
vii)
etc
University of Nairobi –FCE 511 Geotechnical Engineering IV
- 82Spillway
i)
ii)
iii)
iv)
v)
vi)
Introduction to dam design
You might wish to recalculate the adequacy of the spillway. This topic is
covered separately under hydrology
Check field indicators of adequacy of the spillway – water marks
Blockages
Is the gear control working
Structural failures in the concrete
Note any cracks
Gauge house
i)
Are the instruments in good working order
ii)
Have they been vandalized
Reservoir area
i)
Assess the siltation
ii)
Assess the conservation measures being undertaken in the neighborhood of the
dam
iii)
What is the state of the fence of the reservoir fence for the fenced reservoirs?
AOB
3.4
Examples of earth dams in Kenya
In general the dam axis should be chosen in such a way as the material required for the
embankment is minimal while getting the maximum storage. Usually this is so where the
contours are narrowing downstream of a wide valley. The dam axis should be designed as
straight as possible unless the topographical features dictate otherwise
The height of the embankment should be determined in order to achieve the desired
storage with an increased gross freeboard. The gross freeboard is the height between the
spillway crest and the embankment crest and takes account of the design flood and the wave
height
The crest width should be such that earthmoving equipment can be able to work on
the crest. In many cases a road should cap the embankment. In any case a minimum width
of four meters should be observed.
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Introduction to dam design
Slope protection
4 Core
Depth
(m)
Side slopes
Core slope protection
5. Foundations
Soil type
6. Reservoir
Fetch
(m)
Depth
(m)
Area
(m2)
Capacity
(Mm3)
7. Bellmouth Spillway
Crest diameter (m)
Shaft diameter (m)
capacity
(m3/s)
8. Draw off system
Height of stand pipe(m)
Pipe Diameter
(mm)
Height of tower (m)
Table 3.2: Design statistics for Ndakaini dam
1Thika
Description
district
Dam name
Ndakaini (Thika)
District
Thika
Dam type
Zoned embankment
Designer/Engineer
Howard
&
1. Catchment area
Humpreys
Catchment
area 71
(km2Altitude
at dam site 2000
)
(masl)
Mean annual rainfall 1500
(mm)
General soil types
Grade IV to VI
3.
Embankment
Crest length
(m)
420
Crest height
(m)
65
Crest width
11
Bottom
width
(m)
(m)
Upstream side slope
3:1
Dow stream side slope
2.5:1
Freeboard
2
Embankment
(m)
2.5
volume(Mm3)
Crest
2000 masl
10000
culvert
intake
3
1
1
2.75
1
2
1:1.5u/s ,1:5d/s
Filter drains
Weathered
rock
4250
41
2900000
70
15
2
417
20
5500
70
2045 masl
11000
2.25
2025 masl
1
Downstreamshoulder
4000
5
Core
Drainage layers
2.25
1
2005 masl
1
70000
Full storage level 2041 masl
2.5
2030 masl
1
3000
Draw-off tower
3
1
2015 masl
1
1.5
3000
upstreamshoulder
3.5
Riprap
4000
Filter drain
2.25
Drainage blanket
1985
masl
1
5000
2.2
Original Ground level
Draw-off pipe
70 mdeep grout curtain
(a): Embankment details
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1
outlet
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Introduction to dam design
16mdiameter belmouth
embankment
5.5mpipe
(b) bellmouth spillway
Figure 3.2(a): Embankment details of Ndakaini Dam
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Introduction to dam design
Table 3.3: Design statistics for Kwa Tabitha dam, Kitui district
1. Description
Slope protection
Dam name
District
Dam purpose
Designer/Engineer
4
Core
Width
(m)
Kitui
5
Depth
(m)
Domestic water
supply
NWCPC
1
Side slopes
Kwa Tabitha
1. Catchment area
Altitude at dam site 1094
(masl)
Mean annual rainfall 720
(mm)
General soil types
Not available
3.
Embankment
Crest height
Crest width
(m)
Bottom
width
(m)
Hand
riprap
10.5
Core trench
(m3)
5. Foundations
1:2 u/s, 1:2d/s
volume 200
General soil type
Rock
6. Reservoir
Depth
(m)
7.1
7.
Spillway
5
Width at sill
(m)
38
15
Upstream side slope
Depth
(m)
3:1
2
Dow stream side slope
2:1
Freeboard
(m)
Embankment
1.5
12000
(m)
volume(m3)
25000
8.
Excavation
3
(m
) off system
Draw
Height of stand pipe(m)
10
Pipe Diameter
200
Crest
2300
500
1500
Gross freeboard
1
Drawoff
pipe
Protective
Hand placed gravel 300mm
riprap300mm
Grassing
Core
2
1
sand filter
Toe drain
5000
Figure 3.3: Embankment details of Kwa Tabitha Dam, Kitui district Dam
University of Nairobi –FCE 511 Geotechnical Engineering IV
(mm)
1101 masl
crest protection murram
1000
10mstand off pipe
5000
Normal water level ( 1100 masl)
3
placed
5000
1000
Cattle
trough
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Introduction to dam design
Table 3.4: Design statistics for Birica dam, Nyeri district
1 Description
4 Core
Dam name
Trench Width
Birica
District
Depth
Nyeri
Dam type
Side slopes
Embankment
Designer/Engineer
Core trench
NWCPC
(m3)
1. Catchment area
5. Foundations
Altitude at dam site
(masl)
Mean annual rainfall
(mm)
General soil types
3.
Embankment
Crest length
(m)
Crest height
(m)
Crest width
(m)
Bottom
width
(m)
Upstream side slope
Dow stream side slope
Freeboard
(m)
Embankment
3
)
volume(m
Slope protection
2161
1500
Not available
138
7
5
44
2.5:1
2.5:1
1.5
22500
Hand
riprap
1500
5000
placed
Rock
4
22000
50000
15
2
1800
2
200
2169 masl
5000
300mmProtective gravel
Normal water level
2.5
1
2.5
300mmHand placed riprap
1
Original ground level
2mlong stand-off pipe
7000
volume
Soil type
C’
6. Reservoir
Depth
(m)
Area
(m2)
Capacity
(m3)
7. Spillway
Width at sill
(m)
Depth
(m)
Excavation
(m3) off system
8. Draw
Height of stand pipe(m)
Pipe Diameter
(mm)
Crest
Gross freeboard
(m)
(m)
sand filter
8000
15000
Figure 3.4: Embankment details of Birica Dam, Nyeri district Dam
University of Nairobi –FCE 511 Geotechnical Engineering IV
8000
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Introduction to dam design
Table 3.5: Design statistics for Kwa Kasenga dam, Machakos district
1. Description
4 Core
Dam name
District
Kwa
Kasenga
Dam
Machakos
Trench Width
(m)
5
Depth
(m)
1
Dam type
Embankment
Side slopes
Designer/Engineer
NWCPC
1. Catchment area
2. Catchment
2
(km
)
3. Altitude
at
area 360000
dam site
1:2 u/s, 1:2d/s
Core trench
3
)
5. (m
Reservoir
volume 600
Fetch
(m)
Depth
(m)
(masl)
4. Mean annual rainfall 900
(mm)
5. General soil types
Not available
a.
Embankment
Crest length
(m)
Crest height
(m)
Crest width
(m)
Bottom
width
Area
(m )
Capacity
6.
Spillway
120
7
7
Length
41
Upstream side slope
3:1
7.
Excavation
3
(m
) off system
Draw
Dow stream side slope
2.5:1
Height of stand pipe(m)
2:1
Pipe Diameter
(m)
Freeboard
(m)
Embankment
3
volume(m
)
Slope protection
4.5
2
15000
3
(m )
7018
Width at sill
(m)
12.5
Depth
(m)
2
(m)
Not available
1800
Not available
(mm)
18000
Hand
riprap
placed
Crest
7000
Gross freeboard
1500
500
Normal water level
5000
3
1
300mm Protective gravel
Homogeneous embankment
300mm Hand placed riprap
2.5
7000
1
1000
sand filter
4000
5000
1000
Figure 3.5: Embankment details of Kwa Kasenga Dam, Machakos district
University of Nairobi –FCE 511 Geotechnical Engineering IV
Chapter Four : Site Investigation
4.1
Introduction
Site investigations are also referred to as soil exploration. It consists of investigating the
condition on which construction is planned. From site investigation it should be possible to
obtain information for the following geotechnical engineering activities
i.
ii.
iii.
iv.
v.
Design of new foundations
Modification of existing foundations
Location of materials of construction of roads, runways, etc
Identification of materials needed for the construction of pavement structures for
roads, runways etc
Identification of ground to be excavated in the construction of various facilities
including water pipe lines, building foundations, earthworks in cut areas etc
The site investigation should form a part of a coordinated chain of design from inception of
the project through preliminary to the final detailed design of a civil engineering project. It
should indeed continue post construction monitoring of the completed schemes. Because of
the diversity of civil engineering schemes a set of standard procedures is not possible for all
site investigations. The varying civil engineering schemes require a variety of options in
breadth and detail needed for the various schemes. The objectives for which a site
investigation is carried out also differ with various schemes. The main objectives of carrying
out a site investigation are now presented
i) Suitability of site for particular works
In the case of option of site for particular works a detailed site investigation should be able to
enable determination of the most suitable site. Thus it is possible to shift a bridge from one
location which would call for expensive deep foundations to one where ordinary shallow
foundations would be sufficient.
ii) Adequate and economic design
A site investigation leads to safe structures during and after construction. Additionally
sufficient information is obtained for quantifying the excavations needed in the preparation of
the bills of quantities. This should minimizes the possibility of cost overruns due to
unexpected ground conditions being met at construction time.
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Site Investigation
iii) Planning construction
By identifying different materials along the construction paths and their locations a
systematic procedure of carrying out the works is evolved. In the case of road works
materials from the cut areas are analyzed for use in the fill areas. It is then possible to
proceed with construction of the fills and cuts methodically with minimum haulages and
waste of materials.
iv) Prediction in changes in structure
Carefully and well executed site investigations should enable the prediction of the likely
settlement of structures under construction. Equally important is the ability to predict the
effect of excavations on the neighboring structures.
v) Safe structural design of large structures
Heavy modern structures require more detailed site investigations. Today we are seeing
higher buildings, larger bridges and installations sensitive to settlement. Structures and civil
engineering schemes are being put up very quickly. Immediate and consolidated settlement
is taking place when the works are commissioned. Further settlement takes place during the
useful life of the civil engineering installation. Accurate estimation of the settlement regime
is particularly important considering that clients are becoming more and more sensitive to the
performance of structures and the argument that cracks are minor and do not pose any
danger to the structure is no longer good.
4.1.2 Planning a site investigation
Table 4.1 shows a schematic way in which various activities with respect to site investigation
can be performed at various stages of a project. It is clear from the table that site
investigation should not be treated as an afterthought but rather should grow with the project
from conceptual initial design to eventual post construction period.
University of Nairobi –FCE 511 Geotechnical Engineering IV
Table 4.1 Stages of a site investigation
Phase
Stage
Main activity
Site
investigation
activity
SI Reports
Pre-construction
Conceptual
Initial design
Conceptual
design
Preliminary design
Design Alternatives
Detailed design
Detailed Site
Detailed investigations
Desk study of SI – -Boreholes
Review of existing -Trial pits etc
Define Scope of data Preliminary trial Laboratory and field
SI
pits
tests
Terms
of
reference
and i)
Preliminary SI
bid documents
investigation report
Detailed design report
ii) Cost estimate of SI -SI report
Construction
Supervision
construction
Post Construction
of Operation
Maintenance
Construction control
&
Field observations
– field densities
- field moisture contents -
-Performance
Monitoring
and
checking performance –
- pore water pressures
Settlement
Inclinations
As built SI report -
-Maintenance reports
-Performance reports
-Research reports
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4.2
Preliminary and detailed stage site investigations
4.2.1 Preliminary stage site investigations
This should lead to information needed for the design of the various alternatives at the
preliminary stage of the study. The activities in this stage can be summarized as follows:
i)
ii)
iii)
iv)
A study of any existing site investigation reports for the area or in the neighborhood
should form the basis of this stage of investigations.
A study of geographical a geological maps of the site in the case of large sites.
Topographical characteristics should lead to useful information such faulty areas.
Heavily forested areas are an indication of deep rooted top soils.
A site inspection of the existing buildings and any existing structures. Any signs of
distress which can be related to the settlement of the foundations. Any information
from archives, previous records held by the local authorities.
Inspection of the soil profiles, in cut areas, old used quarries. Structured questions to
local people with regard to the geotechnical information being sought yields
considerable information. Such questions are:
a)
b)
c)
What is the depth of the pit latrines in the area?
At what depth murram encountered?
At what depth was water struck?
v)
Aerial survey of the site could give useful information with regard to land formations
and soil profiles.
vi) Seismic refractions could be carried out at this stage of investigations. Usually a
specialist is needed to interpret the results.
vii) Preliminary trial pits
Geophysical methods
Geophysical methods involve sending of seismic or electrical waves through the ground. The
determination of the soil strata is based on the fact that the velocity or the resistance seismic
wave transmission or resistance to electrical flow differs with different rock types and soils.
The method allows the boundaries of the soils to be determined seismic refraction is
described below
Seismic refraction is conducted by having a source of seismic waves (Figure 4.1). The
seismic waves are induced by detonating a small explosive or by striking a metal plate hard.
Waves are subsequently emitted in all directions, through the air, and through the soil in all
directions. Seismic wave transducers called geophones are placed radially from the
epicenter. A circuit connects the geophones and the detonator for accurate determination of
time. A direct wave will reach the geophone first since it is the shortest distance covered.
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Site Investigation
When there is a dense stratum at depth a refracted wave will travel along the top of the bed
rock. As it travels it leaks energy to the surface which can be picked by the geophone.
Seismic source
Geophones
Figure 4.1 Seismic refraction – arrangement of equipment
Time
For short distances the direct waves reach the geophones first. For longer distances
the refracted wave reaches first though the distances is longer than t he surface direct
distance. This is so because the speed of the wave in the dense material is higher than that in
the overburden material of less density. The geophone has a mechanism which records the
first wave and ignores the others. This enables a plot of arrival time versus the distance.
The first section of the graph represents the direct wave measurements while the second
section represents the refracted wave measurements (
Figure 4.2). The inverse of these curves are the velocities of the seismic waves. The
general types of the rocks are determined by geophysics from the knowledge of velocity
versus rock type. It is also used in the determination of depth to water table and thicknesses
of multiple strata. The depth D to the bedrock can be estimated from the formula.
d
Distance
Figure 4.2 Time versus distance for seismic waves
√
4.2.2 Detailed stage site investigations
At this stage the aim is to obtain data for use in the final design of the works. The
investigation is carried out by use of trial pits, sounding and boring. The extent of the use of
these methods depends on the type of the project at hand and the geotechnical parameters
being sought.
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Site Investigation
The trial pits
The pit and shaft technique supplies the most detailed and reliable data on he existing soil
conditions. Once the trial pit has been dug stratification of the soil should be done usually in
the field. In addition as much information should be recorded. This information includes
i.
ii.
iii.
iv.
Depth to ground water table.
Field assessment of the bearing capacity.
Depth of the various strata encountered in the trial pit.
The encountered soils should classified by visual inspection
a.
b.
c.
v.
Coarse grained soils should be described with adjectives such as angular,
rounded with traces of fines etc
Fine grained soils should be studied to indicate whether they are loamy, of low
plasticity, whether they are sandy clays etc
All soils should be described indicating their color and odour if any. Decaying
organic matter if encountered should be mentioned.
Obtain undisturbed samples when you can for the different layers of strata
encountered. These samples can then be taken to the laboratory for tests
For large sites the pits should then be surveyed and located in a grid system for incorporation
into the site investigation report.
Sounding tests
These are basically are penetration tests carried out to supplementing trial pits and borings.
The penetration resistance is measured and related to the bearing capacity. They are widely
used in site investigations. They consist of the cone penetrometer already presented in
chapter 1. The other commonly used penetration equipment is the dynamic cone
penetrometer used in the estimation of the California bearing ratio (CBR) of road pavement
layers. This enables the design of the pavement layers to be carried out
Boring methods
When a deep stratum has to be investigated it will usually be necessary to perform boring
operations to ascertain the strata below the ground to be used in the support of the proposed
structures. Several boring methods are available and are summarized as follows
Percussion drilling consists of a derrick, a power unit and a winch carrying a light steel cable
which passes thorough a pulley. The unit can be towed by a vehicle after the assembly is
folded. The assembly drops a chisel on the ground and strata being drilled
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Site Investigation
Rod
Chisel
Figure 4.3 Schematic presentation of a drilling chisel
The excavation is effected by the drilling chisel. The drilling rods provide the necessary
weight for the penetration the strata. Further weight may be added when need arises. The
winch raises and lowers the chisel and its attachments
Below the water table the loosened soil forms slurry. Above the water table water is
introduced to form the slurry. Periodically the slurry is bailed out by a shell or a bailer to
make progress into the soil. In boreholes which are liable to collapse the borehole must be
cased. In some cases the casings slide on their own weight. On completion of the job, the
casing is jacked out.
Percussion drilling is usually done in diameters of 150mm to 300mm. the borehole
depth investigated by this drilling method can be up to 50 to 60 metres. This method of
drilling can be done on virtually all types of soils including those with boulders and cobbles.
The rig is versatile enough to place mechanical augers and penetrating testing equipments at
appropriate depths.
Power operated augers are usually on vehicles. Downward pressure is applied by pressure
or dead weight. The augurs are 75-300mm diameters. Augers are usually used in self
supporting soils. Casing is usually not needed since the augers have to be removed before
driving. In full flight augers the rod and the helix cover the entire length being investigated.
The augur is then brought up. The soil is ejected by reverse rotation. The likely hood of soil
from different strata being mixed up is very high. In the short flight augur the auger is
advanced into the soil and then raised. The soil is also ejected by reverse rotation.
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Full flight augur
Site Investigation
Short flight augur
Figure 4.4 full flight and short flight augurs
The continuous flight augurs are sometimes fitted with a hollow stem which is plugged
during the drilling operations. When samples are needed the plug and the rods are removed
and a sampler is introduced for the recovery of a sample. The sample may be undisturbed
depending on the sampler utilized. The flight augurs are not suitable for use in loose soils
which are likely to collapse as the augur is inserted and removed from the hole.
Hand and portable augers are usually operated by persons by turning the handle of the
augur. The hand augers are typically of 75 – 300mm diameters. The soil is locked in the
auger and frequent removal is needed to ensure that the augur does not get stack in the soil.
Undisturbed samples may be obtained by introduction of small diameter tubes which are
hammered into the strata under investigation. This method is suitable for self supporting
soils. It is not possible to penetrate coarse granular soils.
Figure 4.5 schematic representation of a hand augur
Wash boring is a method of boring where water is pumped through boring rods and released
through narrow holes in the chisel attached at eth lower ends of the boring arrangement
(Figure ****).
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Water from pump
Tiller
To sump
Drilling bit
Figure 4.6 schematic representation of a hand augur
In this method the soil is loosened and broken by water jet. This is aided by the up an down
movements of the chisel. An attachment to the rods called a tiller enable the rotation on the
drilling bit. The drilling winch is able to raise and lower the chisel and hence get the
chopping action of the chisel.
This method is suitable for most soils but progress is slow if the particles of coarse
gravel larger particles are present. The accurate identification of the soil types is difficult.
The method cannot be used to recover soil samples for testing. However tube samplers can
be advanced into the borehole for obtaining relatively undisturbed samples.
Rotary drilling is done by use of drilling bits that cuts and grinds the subsoil or rock at the
bottom of the borehole. Water is usually pumped down hollow rods passing under pressure
through to the drilling tools. This cools and lubricates the bits. The fluid also provides
support for the borehole where there is no casing.
Two methods of rotary drilling are available. The first is open drilling where the soils
and rocks are broken within the diameter of the hole. Subsequently the tubes are removed and
tube samplers and testing continues below the borehole. This advances the drilling. The
second method is known as core drilling and involves creation of an annular hole in the
material and intact rock enters the drilling core. This advances the drilling and enables
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Site Investigation
samples to be retrieved from the borehole. The sample is then subjected to immediate field
description and taken to the laboratory for various tests. Typical core diameters range from
41mm to 165mm. The method is fast, but in large gravelly soils the speed is slowed by
rotation of the bit without advancement into the ground.
4.2.3 Sampling
Disturbed samples
Disturbed samples are recovered from trial pits and along drilling tools where there is no
attempt to retain the soil constituents. Disturbed samples should however be collected
carefully and placed in airtight tins or jars or in plastic sampling bags. The samples should be
labeled to give the borehole or trial pit identification number, depth of recovery and field
description should be done. The disturbed samples are used for identification tests namely
Field moisture content, PI, grading, compaction and CBR.
Undisturbed sample – cohesive soils
Undisturbed samples are recovered from trial pits and along drilling tools where there is an
attempt to retain the soil constituents. Such a sample is taken in an airtight container with
wax at both ends to prevent moisture from escaping during transportation to the laboratory.
In trial pits the samples can be obtained by pressing a sampler into the ground at the
appropriate depth. The sampler is typically 100mm diameter by 150mm long. In the hand
augur a 38mm sampling tube with a length of 200mm is fitted to the rod after the removal of
the augur. The tube is pressed into the soil and given half a turn to break the soil. The
sampler is then removed and the ends are waxed. In boring rigs a 105mm diameter sampler is
introduced to the borehole to recover a 100mm diameter sample. The sample is usually
381mm long and is fitted with a cutting shoe of about 110mm diameter. The sample is driven
by a falling weight. Any entrapped air or water is expelled from the top through a non return
valve. For soft clays thin walled samplers are preferred to minimize disturbance.
Inevitably there will be some disturbance in the process of retrieving soil samples
from the ground. The least disturbance is for shoes samples cut from the floor of trial pits.
Sample tubes, inserted by pressing, jacking or steady hammering produce some form of
disturbance depending on the thickness of the sampler walls. The degree of disturbance is
related to the area ratio of the sampler tube as given by Equation ****** In general good
samplers have and area ratio not exceeding 25%. Area ratios less than 10% are very good
and are used for very sensitive soils.
x100%
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Site Investigation
De
De
Di
Di
Sampler tubes
Sampler tubes fitted
with a cutting shoe
Figure 4.7 Typical sample tubes
Undisturbed sample – cohesionless soils
Various methods have been employed to obtain undisturbed sand samples. These include
freezing, chemical application, and use of compressed air (Smith and Smith, 1998).
Whatever method is employed eventual disturbance occurs as the soil is transported to the
laboratory for testing. In light of these difficulties it is prudent to assess the engineering
properties of cohesionless soils through field testing such as penetration.
Quality class for soil sampling
Table ** below based on Rowe (1972) shows the quality classes for soil samples obtained
from various site investigation operations.
Table 4.2 Quality class for soil sampling
Quality
class
5
4
3
2
1
Method of sampling
Material brought up by drilling tools an no attempt is
made to retain all the soil constituents
As for 5 but all soil constituents are retained as far as
possible. Bulk an jar samples. Plastic bag samples
Pressed or driven thin or thick walled samplers with
water balance in very permeable soils
Use of sample
Rough sequence of
strata
Sequence of strata and
remolded properties
As
above
and
examination of soil
fabric
As for class 3 above but with water balance all the As 3 and γ, n, mv, cu,
time
c’ θ’
Thin walled piston samplers with water balance
As 2 and cv and k
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Site Investigation
Borehole logs
Borehole logs summarizes all the laboratory an field tests carried out on samples representing
the various strata encountered in the boring operations. All ground conditions encountered at
the site are also included. The log enables a rapid accurate assessment of the soil profile on a
vertical scale. The details of the various strata encountered including all their geological
formation details which can be inferred are given. The details captured should include the
depth to which ground water was encountered. The description is based on particle
distribution and plasticity based visual inspection and feel. Soil color should also be
recorded.
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Courtesy of Norken Engineering Consultants
Figure 4.8 Borehole logs
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Site Investigation
4.2.4 Scope of Site Investigation
Spacing of the trial pits and or boreholes
The scope of site investigation is dependent on the effect of the construction on the ground.
The scope should be commensurate with the needed geotechnical parameters. Table 4.3
shows the suggested minimum number of borings for the various structures.
Table 4.3 Recommended spacing of investigation trial pits and boreholes
Project
Multistory
1 to 2 storeys
Bridge piers
abutments
Type of soil/Distance between borings
Uniform
Average
Erratic
45
30
15
60
30
15
and 30
30
15
Minimum no
4
3
1 – 2 per unit
For highways and runways during preliminary design the subgrade soils along the proposed
alignment should be sampled at 1000metres and the samples should be tested to establish the
in-situ CBR, grading and plasticity of the materials. At this stage the material site should be
investigated at 60 meter intervals. In the detailed stage the subgrade is sampled at 500meters
while the material sites are sampled at 30metres.
Depth of investigation
The depth should be such as to capture the geotechnical information needed for the design of
the facility. Equally important is to capture the information needed in the quantification of
the bill of quantities to ensure an accurate specification of the works is carried out. The
recommended depths below the formation of investigation for the various civil engineering
schemes is shown on Table 4.4 based on Figure 4.9 below.
Table 4.4 Depth of investigation
Project
Column foundations
Raft foundations
Bridge piers and
abutments
Earthworks in fill for
highways
Earthworks in cut
highways
Pipe works
Depth
1.5B-3B
1.5B
1.5B-3B
In rock
1.5-3m
1.5-3m
1.5-3m
Parameters to be established
C, θ, N, RQD,TCR
C, θ, N. RQD,TCR
C, θ, N, RQD,TCR
0.5L
0.50m
0.5H
0.50m
D
0
PI, CBR for fill material
Strength of support
Establish the type of excavated
material and strength of support
Investigate type of excavated
material and strength of support
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Site Investigation
Raft foundations
Column
foundations
B
Piled foundations
B
L
B
H
Retaining walls
B
a)
Structural foundations
L
L
In cut H
In fill
b)
Highway earthworks
D
c)
Pipe works
Figure 4.9 Scope of foundation investigations
4.2.5 Site Investigation Reports
List of suitable headings
Title page
Gives the title of the project at a glance
Abstract
The abstract should be approximately 200 words. It is a very important element of the
project and should be prepared with care. It must convey the essence of the site investigation
and all the important findings without ambiguity.
List of contents
Guides the reader to the various chapters
Field work
A brief and complete description of what was done in the field. Boreholes, and trial pits
performed, field testing etc. Actual procedures of standard tests need not be repeated. A
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mention of the tests performed is sufficient. New procedures and peculiar fieldwork should
be explained.
Laboratory work
A brief and complete description of what was done in the laboratory work carried out . as in
the case of field testing actual procedures of standard tests need not be repeated. A mention
of the tests performed is sufficient. New procedures and peculiar laboratory equipment and
procedures should however be explained
Site description and geology
An engineering summary of the nature of the site an its geology, including aspects such
excavated areas and what was found, stability of natural slopes, drainage etc
Engineering properties of soils an rocks
A summary of the results of field and laboratory tests and other observations made at the site
Discussion
A reasoned discussion of what design and construction problems are likely to be encountered
in relation to the site and its geological situations.
Recommendations and conclusions
A brief but clear statement of the recommended geotechnical parameters investigated. The
treatment of the various aspects of design should come out clearly and without doubt. Values
of use in design and construction should be summarized viz, allowable bearing capacity,
estimated settlement, suitable types of foundations, construction requirements namely
grouting, compaction etc
References
A list of the books, papers, referred to in the work
Appendices
Appendix A – should contain site plan, borehole logs, photographs, etc
Appendix B – should contain tables of results of field and laboratory test those not included
in Appendix A
Appendix C – Any special or unusual test procedures adopted in the investigation
References:
Craig FR, 1987, Soil mechanics, Van Nostrand Reinhold (International) London
Bowles JE , 1982, Foundation Engineering, McGraw-Hill international book company,
Tokyo.
Tomlinson MJ and Boorman R (1986), Foundation and construction, Longman scientific and
technical, England
Franklin JA and Dussealt MB (1989) Rock Engineering, McGraw-Hill international editions,
London
Chen FH (1975) Foundations on expansive soils, Elsevier scientific Publishing Company
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